Pore Water Pressure Research Papers

Pile Setup, Dynamic Construction Control, and Load and Resistance Factor Design of Vertically-Loaded Steel H-Piles

Because of the mandate imposed by the Federal Highway Administration (FHWA) on the implementation of Load Resistance Factor Design (LRFD) in all new bridge projects initiated after October 1, 2007, research on developing the LRFD... more

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Effectiveness of horizontal drains for slope stability

by EC Leong

2003, Engineering Geology

Horizontal drains have been commonly used in stabilising unsaturated residual soil slopes. This study examines the effectiveness of horizontal drains in stabilising residual soil slopes against rainfall-induced slope failures under a... more

Fig. 1. Location of NTU-CSE and Nanyang Heights slopes. deposited around the rocky flanks of the island (Leong et al., 2002). sandstone, siltsone, mudstone, shale, tuff, conglom- erate and limestone. The formation has been severely folded and faulted in the past as a result of tectonic movements. The semi-hardened Old Alluvium was deposited by an ancient river system, probably in the Pleistocene epoch, during a low stand of the sea. Residual soils of the granitic and sedimentary rocks occupy about two-thirds of the land area of Singapore. The present day configuration and much of the morphology of the low-lying areas of Singapore is a result of erosion and deposition during the period of fluctuating sea levels in the late Tertiary and Quater- nary. As sea level rose after the end of the last cold stage about 11,000 years ago, it formed Singapore a group of islands, separated from the Malay Peninsular by the straits of Johor. Extensive marine clay, beach deposits and associate terrestrial sediments were The relief of Singapore is relatively gentle. Only 10% of the land area is over 30 m high and more than 60% of the land is less than 15 m high (Pitts, 1984a). The area of granitic and other igneous rocks in the center of the island forms a landscape of rounded hills and gentle spurs and valleys. The coast is flat, but in few places cliffs can be found. Considerable stretches of the coastline have been markedly modi- fied by reclamation work, building of embankments and swamp clearances.

Fig. 4. Soil—water characteristic curve and permeability function for purple soil at NTU-CSE slope: (a) soil—water characteristic curve, (b) permeability function.

Fig. 5. Average pore—water pressure readings at various depths of the NTU-CSE slope: (a) monitoring period: April—June 1996, (b) monitoring period: July-September 1996.

Fig. 7. Pore—water pressure profiles for drained and undrained area at Nanyang Heights slope: (a) drained area, (b) undrained area. the pore—water pressures at all depths followed the same trend, rising and falling at the same time with respect to rainfall and evaporation. For the depths at 0.3, 0.8 and 1.4 m, very little distinction could be made between the pore—water pressures in the undrained and drained areas. At the 2.3- and 3.2-m depths, there was a difference in pore—water pres- sures between the undrained and drained area, which was maintained at about 15 kPa. The validity of the results obtained at 2.3 and 3.2 m in the undrained area is questionable. Proper contact between soil and the ceramic cup of the tensiometer may not have beer achieved and this may explain why these tensiometet readings seemed to be relatively unresponsive when compared with those in the drained area (see Fig. 72 and b).

Fig. 8. Finite element model for the parametric study: (a) slope configuration and horizontal drain location, (b) finite element mesh

A factor of safety was calculated for each drainage scenario at every time step by importing the pore— water pressure head files into Slope/W model. The results of the slope stability analyses are shown in Fig. 10. The general trend of Fig. 10 shows that the least amount of benefit was derived from the drains located in the upper region of the slope. The most benefit was derived from the drain located at the bottom of the slope and a very small additional benefit was obtained by combining the bottom drain with other drains in the upper region. This is shown by comparing the factor of safety obtained from the case with three drains as opposed to the case with one drain at the bottom of the slope. The condition with the three drains only gives a 3% additional factor of safety as compared to the condition with the bottom drain. The reason for the slight difference in factor of safety is that the bottom drain attracted the most water. It can be seen in Fig. 9 that the seepage flow in the bottom zone was relatively the same when either a single drain was present in this bottom zone or if all three drains were present in the slope. When the three Fig. 9. Theoretical flux through bottom zone for various drainage configurations.

Fig. 10. Theoretical factor of safety with respect to drain position for a constant rainfall rate.

USCS: Unified Soil Classification System. Properties of two soil types found at NTU-CSE slope

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Impact and analysis of geotechnical processes on earthfill dam breaching

by Mahmoud al-Riffai and

2010, Natural Hazards

For over three decades, emergency planners have used numerical models to predict breaching in earthfill dams due to extreme events such as overtopping. However, current models neglect the role of the unsaturated zone present within the... more

Fig. 1 Coupling between modeling processes and input parameters in a typical physically based numerical model

Fig. 2 Outflow hydrograph and the proposed four breaching phases (Orendorff et al. 2009)

Fig. 3 Transient response of the phreatic surface during a rapid reservoir drawdown in non-cohesive earthfill dam n classical soil mechanics, soils are considered to be saturated where the pore-air-pressure (PAP), u,, is zero, and PWP, u,,, is positive and increases linearly with the depth below the phreatic surface. However, the downstream face of earthfill dams is unsaturated and the phreatic surface is depressed with the aid of seepage control structures, such as drainage systems or central cores constructed within the earthfill dam. Drainage systems are usually inverse-type filters (i.e., a horizontal or vertical drainage blanket) installed in non-cohesive homogenous earthfills such as dikes. Central cores are constructed of cohesive soils such as clays or clay-silt mixtures. During the rapid drawdown of the reservoir level caused by a breach, there is a transient depression of the phreatic surface (Fig. 3), which further reduces the degree of saturation, S, of the soil. As the head in the reservoir drops, water is drained from the earthfill body and air enters into the inter-granular pores, passing from a saturated state to one in which a three-phase system (solid, water, and air) develops. The air—water interface (also known as the “contractile skin”) is considered to be a fourth phase, which has a considerable effect on the physical behavior, as well as on the prop- erties of partially saturated soils (Fredlund and Morgenstern 1977). Thus, two new stress- state variables govern the behavior and properties of unsaturated soils; the matric suction (or soil tension), u, — u,, and net normal stress, o, — Ug, where u, (40) is the air reference pressure and o,, is the total normal stress on the failure plane. Although this stress analysis was first presented by Fredlund and Morgenstern (1977), it was only during the last decade that several methodologies have used this continuum

a AIRE A Se ee The two key soil parameters that influence earthfil Juctivity, k, and shear strength, t. During soil drying, n the soil, rendering the flow path more tortuous and t he permeability of the soil (Fig. 4b). For finer soils dam breaching are hydraulic con- air bubbles occlude the water phase hus causing a non-linear decrease in , the AEV is much larger and the Jesorption rate is lower. It is important to note that for cohesive soils, the soil structure also olays a role in the SWCC parameters. With respect to the compaction moisture content, dry of optimum soils (flocculated soil structures) has a hig her saturated permeability, k..,, than

three initial breach geometries: V-notch, crater or Dlast-induced, and tapered or naturally induced overtopping initiation (Orendorff et al. 2009). The construction of the earthfills was made in 5-cm-thick layers where compaction was at optimum moisture content (OMC), obtained from previous in situ compaction tests. It is important to note that the in situ compaction curves differed from those obtained using the standard Proctor test. The larger peripheral confinement of hydraulic flumes (i.e., in comparison with Proctor molds) resulted in a higher maximum p, and its corresponding OMC for in situ compaction tests. The drainage system was an under toe-drain with an inverse filter. The coarser material was crushed gravel (Ds) = 4,400 um), and the filter material was silica sand (Ds) = 780 tm). The drainage element satisfied the gradation design of horizontal blankets to prevent migration of fines (USBR 1987). The surface elevation of the reservoir during the breaching process was recorded at four locations upstream the upstream toe of the earthfill dam: 0, 0.3, 2.3 and 25.3 m using WG-50 wave capacitance gauges. A FORTRAN code was developed to process the water surface ele- vation data recorded at each time-step into a discharge-time relationship. Three high- definition cameras were also used to capture the breach morphology at three different angles (upstream, downstream and top). For all tests, prior to breaching, the reservoir was impounded at very slow rates and the piezometric head was recorded at the earthfill foundation level using piezometers. A fourth video of the transient response of the pie- zometric head at the channel bottom during breaching was also captured. Once the water surface level reached the base of the V-notch (7 = 280 mm), the reservoir inflow was stopped in order to establish steady-state seepage. Once this was attained, the filling continued with a constant inflow of 0.49 I/s (Fig. 5).

Fig. 6 Comparison between the breach hydrograph for the dam with drain and without drain

Fig. 7 Measured and computed (SEEP/W) steady-state piezometric head distribution along dam foundation at n = 280, 185, 50 mm (without drain—left graph, with drain—right graph) all Using the SWCC and the saturated hydraulic conductivity measured for various voi ratios, the seepage through the dam was modeled using SEEP/W (GEO-SLOPE 2004). Th numerical simulation was calibrated by adjusting the anisotropy ratio of k,. with respect t the horizontal and vertical axes, such that the numerical output of the piezometric head ; the earthfill foundation is in correlation with measurements collected using the piezometet (Fig. 7). PWP distribution plots reveal that the suction in the soil did not exceed 1.5 kPa i the case of a toe-drain (Fig. 8). Attempts at modeling the breaching process using NWS BREACH and HR-BREACH for all of the tests mentioned earlier have been mad However, the output of both models yielded a constant outflow through the breach channe sven for exceeded simulation times, implying that no erosion took place. For the smaller-scale tests conducted in the narrower flume, the breaching was to

Fig. 8 Steady-state PWP distribution [kPa] and phreatic surface (without drain—left plot, with drain—right nipnty)

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The effect of antecedent rainfall on slope stability

by EC Leong

2001, Geotechnical and Geological …

A case study is presented in order to identify the effect of antecedent rainfall on slope stability for Singapore. A storm in February 1995 (during which 95 mm of rain fell in 2 1 2 h) caused more than twenty shallow landslides on the... more

Figure I. Locations of landslides on the NTU campus due to the storm of 26 February 1995

Figure 2. Daily rainfall for the months of (a) November 1993 (b) November 1994 (c) February 1995

Figure 3. Relationship between antecedent rainfall and the occurrence of landslide:

Figure 5. Soil-water characteristic curve for a residual soil from the Jurong Sedimentary Formatior

Figure 6. Permeability function for a residual soil from the Jurong Sedimentary Formation

Figure 7. Rainfall intensities with elapsed time used in simulations for (a) Case | (b) Case 2 (c) Case :

Figure 8. Simulated pore-water pressure contours for Case | at elapsed times of (a) 0 h (b) 120 |

Figure 8. Simulated pore-water pressure contours for Case | at elapsed times of (c) 125 h (d) 144 h

Figure 9. Simulated pore-water pressure profiles with depth for Case _

Figure 10. Factor of safety variations with elapsed time for Case |

Figure Il. The critical slip surface for Case 1 (elapsed time = 125 h)

Figure 12. Simulated pore-water pressure contours for Case 2 at elapsed times of (a) 0 h (b) 120 h

Figure 12. Simulated pore-water pressure contours for Case 2 at elapsed times of (c) 122.5 h (d) 144 |

Figure 13. Simulated pore-water pressure profiles with depth for Case 2

Figure 13. Simulated pore-water pressure profiles with depth for Case 2
Figure 14. Factor of safety variations with elapsed time for Case 2

Figure 15. The critical slip surface for Case 2 (elapsed time = 122.5 h)

Figure 16. Simulated pore-water pressure contours for Case 3 at elapsed times for (a) 0 h (b) 120 |}

Figure 16. Simulated pore-water pressure contours for Case 3 at elapsed times for (c) 122.5 h (d) 144 1

Figure 17. Simulated pore-water pressure profiles with depth for Case 3
Figure 18. Factor of safety variations with elapsed time for Case 3

Figure 17. Simulated pore-water pressure profiles with depth for Case 3

Figure 19. The critical slip surface for Case 3 (elapsed time = 122.5 h)

Figure 19. The critical slip surface for Case 3 (elapsed time = 122.5 h)
Figure 20. Comparison of infiltration to rainfall rates for Case 3

Table I. Details of landslides on the NTU campus due to the storm of 26 February 1995

Table 1. (continued)
*Indicates estimated values

Table 2. Basic properties of the residual soils

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Calciclastic submarine fans: An integrated overview

by Aitor Payros

2008, Earth-Science Reviews

Calciclastic submarine fans are rare in the stratigraphic record and no bona fide present-day analogue has been described to date. Possibly because of that, and although calciclastic submarine fans have long intrigued deep-water carbonate... more

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Estimation of Degree of Consolidation for Vacuum Preloading Projects

by Tùng Hoàng

The degree of consolidation is usually used as one of the criteria for assessing the effectiveness of soil improvement work using the fill surcharge or vacuum preloading method. It is also often used as a design specification in a soil... more

Fig. 1. Schematic illustration of pore water pressure and effective stress changes under (a) fill surcharge and (b) vacuum load

The method shown in Eq. (1) has the following advantages over the method using settlement data: (1) The DOC calculated using Eq. (1) relies only on field pore water pressure data, whereas when calculating the DOC using settlement data, the ultimate settlement has to be predicted; (2) Not only the final DOC, but also the DOC at any time can be calculated using Eq. (1), as u,(z) represents the pore water pressure at any time, t; and (3). For consolidation involving multiple layers, Eq. (1) can be applied to any single layer to calculate the DOC achieved in a particular layer. In this case, the upper and lower limits of the integrals in Eq. (1) are set to be the top and bottom of that soil layer. However, it is not easy to calculate the DOC for each layer for multilayer soils using settlement, as the settlement of each In Eq. (1), uo(z)=initial pore water pressure at depth z; u,(z) =pore water pressure at depth z at time t; u,(z) =suction line, z =depth; y,,=unit weight of water; and s=suction applied. The

Fig. 4. Project site and plan view of instrumentation for Case I

Fig. 3. Schematic illustration of pore water pressure distributions versus depth under both combined surcharge and vacuum load

Fig. 7. Pore water pressure reductions plotted against duration of vacuum loading A field instrumentation and monitoring program was imple- mented in this project. Instruments including surface settlement plates, deep settlement gauges and pore water transducers were used as shown in Figs. 4 and 5. Data were recorded at frequent

Fig. 6. Settlement measured at different depths plotted against duration of vacuum loading

Fig. 8. Pore water pressure distributions at different durations for Section I

Fig. 9. Project site and plan view of instrumentation for Case II Under vacuum load, the settlement and pore water pressure reductions measured at different depths at Section I are plotted against duration in Figs. 6 and 7, respectively. It can be seen from Fig. 6 that the ground had settled nearly 900 mm toward the end of consolidation. Fig. 7 shows that at the end of vacuum loading, the pore water pressures had reduced by 70-140 kPa. Using the data presented in Fig. 7, the initial and final pore water pressure distributions together with the hydrostatic pore water pressure line and the suction line are plotted versus depth in Fig. 8. The initial pore water pressures were greater than the hydrostatic pore water pressure as the subsoil was still undergoing consolidation. The pore water pressure distributions at 30, 60, and 90 days are also plotted in Fig. 8 to illustrate the pore water pressure dissipa- tion process. It can be seen that the final pore water pressure distribution is very close to the suction line. This is indicative that the vacuum preloading method was very effective throughout the entire soft clay. Applying Eq. (1) to Fig. 8, the DOC calculated at the end of the project was 95%.

Fig. 11. Settlement measured at different depths plotted against duration of vacuum loading at: (a) Section I and (b) Section II

Fig. 10. Applied vacuum pressure and ground settlement measured with duration

Fig. 12. Pore water pressure reductions measured at different depth plotted against duration of vacuum loading at: (a) Section I and (b) Section II

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The effect of antecedent rainfall on slope stability

by David Harianto

2001, Geotechnical and Geological Engineering

A case study is presented in order to identify the effect of antecedent rainfall on slope stability for Singapore. A storm in February 1995 (during which 95 mm of rain fell in 2 1 2 h) caused more than twenty shallow landslides on the... more

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Using Stone Column as a Suitable Liquefaction Remediation in Persian

by Mehdi Mokhberi

This research investigates behavior of gravel drain piles under high-level earthquake loading beneath the structures foundation. To achieve this purpose one of the waste water septic tank project in north of Persian Gulf in Hormoz Island... more

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GIS-based soil liquefaction susceptibility map of Mumbai city for earthquake events

by Sumedh Mhaske and

2010, Journal of Applied Geophysics

Fig. 3. Mumbai and its suburban area showing marshy and muddy land along sea shore. where (N;)¢o,cs is the corrected SPT value equivalent to clean sand condition; factors a and 6 are computed as,

Mumbai and its suburban area are taken as the study area for evaluation of liquefaction potential of soil. A total of 238 numbers of borehole data, both undisturbed and disturbed soil test reports and rock core test report were collected from various Government engineering institutes and professionals working in geotechnical engineering field. After studying these data it is found that all the soil parameter related to analysis of evaluation of liquefaction potential of soil are sometimes insufficient. Hence from available field data, the required soil para- meters are obtained from standard charts given by Murthy (1992). The soil parameters given in Table 3 are used for determining SPT value ‘N’ and relative density ‘D,’ for given value of soil friction angle . The data given in Table 4 are considered to determine the corrected ‘N’ value, total effective stress, CRR value and the factor of safety against liquefaction of soil. Factor of safety against liquefaction of soil (FS,) at different earthquake moment magnitudes is determined by using field data of study area as given in Table 5. The analysis is done for different earthquake moment magnitudes My = 5.0 to My = 7.5. Comparisons of results of the factor of safety values against liquefaction of soil are given in Table 6. It can be seen from Table 6 that except for Kadamwadi marol Andheri (East) site, all other sites in Mumbai city considered for present study are subjected to relatively low value of factor of safety against liquefaction (FS,). The reason for getting a high value of FS, for Kadamwadi marol Andheri (East) site can be justified as the observed SPT ‘N’ value is very high for this site compared to the other sites and Plate 1. Google map showing area of Chikuwadi, Borivali which is prone to moderate liquefaction during earthquake moment magnitude My = 7.0 to 7.5

Reclamation of land around the seven islands of Mumbai (Kaur, 2001). Table 2

Past tremors that shook Mumbai city (Wenzel et al., 2007; Joshi, 2007). Table 1

Relationship between N and ¢ to relative density of cohesionless soils (Murthy, 1992). Table 3 Table 4

Input data used for determination of factor of safety against liquefaction of soil.

The bold letter indicates the factor of safety against the soil liquefaction less than 1.3. Comparisons of results of factor of safety against liquefaction of soil.

The summary of type of liquefaction of soil at various earthquake moment magnitudes. Table 8

Details of undisturbed soil data used for evaluation of liquefaction potential of soil.

Classifications for different types of liquefaction of soil during earthquake. Table 7

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Numerical simulation of soil–water interaction using smoothed particle hydrodynamics (SPH) method

by Ha Bui

2007, Journal of Terramechanics

An application of smoothed particle hydrodynamics (SPH) to simulation of soil-water interaction is presented. In this calculation, water is modeled as a viscous fluid with week compressibility and soil is modeled as an elastic-perfectly... more

Fig. 3. Initial arrangement of simulation test. In order to evaluate our SPH model for soil—water inter- action, simulations of excavation by a water jet have been carried out for dry soil and saturated soil. The dry soil is modeled by one type of particles with uniform material properties filled in a rectangular reservoir 50 cm in length

Fig. 1. (a) Particle approximations using particles within the influence domain of the smoothing function W for particle 7. The influence domain is circular with a radius of xh. (b) The cubic spline kernel and its first derivative. with « is a constant related to the smoothing function for point at x and defines the influence domain of the “kernel estimate” (see Fig. 1a). In the SPH method, the calculation domain is represented by a finite number of points (particles), which carry mass

The final equation needed to solve the above Navier— Stokes equations for water is the “equation of state”’ which is used to estimate the pressure change of water. Monaghan [5] applied the following form when he first simulated the free surface flows of water using SPH,

where the subscripts i and j represent for soil particles while a and 5 are used for water particles. It is clear that these momentum equations of saturated soil are different from that of single phase, Eq. (22). The presence of the seepage force in Eqs. (36), (37), and the contribution of pressure from water to soil in Eq. (36) make them possible to simu- late the effect of seepage force and pore water pressure in the saturated soil model, as a result the interaction between soil and water could be simulated through SPH. Fig. 2. Modeling of soil—water interaction in SPH simulation.

Soil phase In the above equations, c and v represent the speed of sound and particle velocity vector; x is the position vector of particle; and ¢ is usually taken as 0.01 ([5]). The param- eter « produces a bulk viscosity while f suppresses particle interpenetration. Monaghan chose « = 0.01, 6 = 0 to simu- ate the free surface flow, Libersky [4] chose «= 2.5, fh =2.5 for material strength simulation. However, neither is suitable for our simulations since we simulate the water jet flow with high velocity and the soil behavior is different from that of solid. This paper we choose « = 0.01, 6 = 1 for water and « = 1, 6 = 1 for soil. The momentum equations for saturated soil after introducing the artificial viscosity to the pressure term are: To model the solid boundary condition, we simply fol- low the method proposed by Monaghan [5]. This method assumed the presence of a ghost line of particles located right on the solid boundary to produce a highly repulsive force to the real particles near the boundary, and thus to prevent these particles from penetrating the solid bound- ary. The repulsive force is calculated similarly to the Len- nard-Jones molecular force as,

Fig. 4. Erosion during excavation by a water jet on dry soil model by SPH simulation. Water jet has speed of vjep = 25 m/s and 20cm in width. These particles have the following soil model is generated by mixing rial properties with the particles in t cosity of 4 = 10-3 Ns/m?. There are properties: density p—=2.7g/cm*, Young’s modulus E=150 MPa and Poisson’s ratio v =0.3. The saturated wo types of particles hat model water and soil phases into the same soil domain. The soil particles in the saturated soil have the same mate- he above dry soil, and he water particles have density of p = 1.0 g/cm? and vis- a total of 4000 parti- cles generated regularly on each phase with an initial Fig. 4 shows the erosion process of soil excavation by water jet on the dry soil model at representative times. In this simulation, both water pressure and seepage force, which results from the relative velocity between water par-

Fig. 5. Erosion during excavation by a water jet on saturated soil model by SPH simulation. Water jet has speed of vj-, = 25 m/

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Liquefaction of Silts and Silt-Clay Mixtures

by Shamsher Prakash

1999, Journal of Geotechnical and Geoenvironmental Engineering

Low plasticity silts and silty clays occur extensively in the Central United States, India and China. For evaluating their liquefaction potential during an earthquake, no accepted guidelines are available based on their density, void... more

Standard Penetration Test Blow Count (N,)¢o Figure 1 Relationship between Stress Ratio Causing Liquefaction and (N1)¢0 values for Silty Sand for M = 7.5 (after Seed et al. 1985)

Figure 2. Chinese Criteria Adapted to ASTM Definitions of Soil Properties (Perlea, Koester and Prakash, 1999)

Figure 3 Variation of Cyclic Strength with Fine Content at Constant Void Ratio (after Troncoso, 1990) Troncoso (1990) compared the cyclic strength of tailing sands with different silt contents ranging from 0 to 30 % at a constant void ratio of 0.85 (Fig. 3). It was found that the cyclic strength decreased with the increase of fine content. This conclusion is in apparent contradiction with that of Seed et al. (1985). Note, however, that the difference of results between Seed et al. (1985) and Troncoso (1990) is due to the different criteria used in their studies. In Troncoso’s study, the cyclic shear strength of tailing sands was examined at a constant void ratio, while in the study of Seed et al., cyclic stress ratio was investigated at the same SPT values.

Figure 4 Cyclic Stress Ratio for Well-Graded Sand Mixtures, with Index Propertie: and Test C onditions Shown (after Chang 1990)

Figure 5 Normalized cyclic Stress Ratio versus plasticity Index on Undisturbed samples (Data of El Hosri et al 1984)

Figure 6 Rate of Pore Pressure Buid up in Cyclic Triaxial Tests on Undisturbed Samples (After El-Hosri et al. 1984)

Figure 7. Cyclic Stress ratio Versus number of Cycles for Undisturbed Saturated Samples for o3 = 10.0 psi (Puri, 1984)

Figure 8.C yclic Stress ratio Versus number of Cycles For Reconstituted Saturated Samples For 03 = 10.0 psi (Puri, 1984)

Figure 9 Comparison of Cyclic Stress Ratios for Undisturbed and Reconstituted Saturated Samples For Inducing u = o3 Condition in a Given Number of Cycles (Puri, 1984)

Figure 10. Cyclic Stress Ratio Versus Number of Cycles for Reconstituted Saturated Samples for Different PI Values, Inducing 5% D.A Axial Strain (Puri, 1984)

Figure 13 Cyclic Stress Ratio versus Number of Cycles for Low Plasticity Silts, for Inducing 20 % DA Axial Strain Condition at 15 psi Effective C onfining Pressure, for Density= 95.4 - 98.5 pcf and w = 8 % (Sandoval, 1989; Prakash and Sandoval, 1992)

eee ae ae a eee ee a Figure 14 Cyclic Stress Ratio versus Number of Cycles for Low Plasticity Silts for Inducing Initial Liquefaction C ondition at 15 psi Effective C onfining Pressure; PI = 1.7, 2.6, and 3.4, for Density 97.2-99.8 pcf, and w = 8% (Sandoval 1989; Prakash and Sandoval 1992) The authors will , however, will like to point that the critical value of PI = 4, mentioned above is based on the results of limited studies. Puri (1984,1990) observed initial liquefaction for samples of clayey silt with PI~10 for both undisturbed and reconstituted samples. However , initial liquefaction was not observed for reconstituted samples with PI of 15 and 20. Sandoval(1989) only tested samples with PI = 1.7, 2.5 and 3.4 . The index properties of the soils tested by Puri (1984) and Sandoval(1989) also had some differences. The results presented in Fig. 15 may therefore be considered only qualitative in nature . Nevertheless, the results do indicate that the plasticity of the soil plays a very important role in influencing the liquefaction behavior of clayey silt. The critical values of PI, and factors affecting this value need further investigation.

Figure 15 Cyclic Stress Ratio versus Plasticity Index for Silt-Clay Mixtures (CSR Normalized to initial Void Ration e9 = 0.74) (Prakash and Guo, 1999)

Table 1: Properties of Different Low-Plasticity Soil Samples (after Ishihara and Koeski 1989) Note: e, = void ratio after consolidation; CSR = cyclic stress ratio causing 5% strain in 20 cycles.

Table 3: Normalized Test Results for Various Numbers of Cycles Note: CSR normalized to initial void ratio e9 = 0.644

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Comparative analysis of contributing parameters for rainfall-triggered landslides in the Lesser Himalaya of Nepal

by Netra Bhandary

2009, Environmental Geology

In the Himalaya, people live in widely spread settlements and suffer more from landslides than from any other type of natural disaster. The intense summer monsoons are the main factor in triggering landslides. However, the relations... more

From the meteorological point of view, the month of July is very critical for the study area. Plotting the last 34 years rainfall data at Thankot (Fig. 3) reveals that month of July of every year has the maximum rainfall. From | to 22 July

Fig. 3 a Trend of monthly rainfall distribution of last 34 years (1970-2004) in study area and the monthly rainfall distribution is in approximately bell-shaped symmetric fashion with peak value in July. b Rainfall distribution of July 2002

Fig. 5 a Oblique view of Matatirtha ridge and location of selected sites M1 and M2 for geotechnical study and seepage analysis. b Oblique view of Dhaksi ridge and selected sites, D1 and D2, for geotechnical study and seepage analysis. View is toward west for both photographs

Fig. 6 a Oblique view of Matatirtha landslide 32 h after landsliding. Seepages through rock were also seen, view from east to west. b Topographical map of landslide

Fig. 7 Grain size distribution of soils of selected sites

Fig. 8 Casagrande plot of Atterberg limits of soils from Matatirtha and Dhaksi ridges. Open triangle represents Atterberg limits of soils from selected site, whereas open circle represents Atterberg limits of soils from surrounding area of selected sites

Fig. 9 Comparison of peak and residual angles of internal friction for representative soil samples from Matatirtha and Dhaksi ridges. The soil samples from sites M1, M2, D1, and D2 are also shown with open circle and open triangle notation

days of July have considerable rainfall and no data was available for pore pressure variation. Thus, the simulation was performed considering daily rainfall intensity begin- ning from Ist July 2002 as a start-up period, and the initial water table was defined above the bedrock with a maxi- mum negative pore pressure of 10 kPa. These kinds of simulations of long start-up periods usually indurate the computed initial conditions (here, pore water pressure) and its effect in the main event is negligible (Reid et al. 1988). imulations of long start-up periods usually indurate the

In modeling with SEEP/W, bedrock hydrology was not incorporated (e.g., Casagli et al. 2006; Dapporto et al. 2005; Crosta and Dal Negro 2003). However, the role of bedrock seepages in slope failures and debris flows is well described in the literature (Mathewson et al. 1990; Onda 2004; Wilson and Dietrich 1987). Therefore, to ascertain the applicability of bedrock seepage in modeling, site M1 (crown of Matatirtha landslide) was simulated with both scenarios of seepage through rock (M1 simulation) and no seepage through rock (M1NFB simulation), i.e., as a nul flux boundary on soil-bedrock contact. For bedrock seep- age, only the extreme rainfall period was considered as effective rainfall for groundwater flow. From the field observations of slopes from 2002 to 2006, the norma monsoon rainfalls (less than 100 mm per day) were not found to be effective for groundwater flow in the uphil section of the study area. However, such rainfalls were responsible for springs and seepages originating at the base section of the study area. However, such rainfalls were

Fig. 13 Pore water pressure variation in July at the optimized slip surface having lowest factor of safety. The optimized slip surfaces were obtained for the day of maximum rainfall and are shown in Fig. 14 for each site The value of pore water pressure in each slope was usually high in 23 July, which was the date of failure of the Matatirtha landslide. The role of antecedent rainfall was almost negligible in all simulations. Pore water pressure development was usually transient because of the high permeability of soil and potential seepage faces on the slope. Pore water pressure variations in July at the opti- mized slip surface having lowest value of factor of safety are shown in Fig. 13. The optimized slip surfaces of each site are shown in Fig. 14 on the day of maximum rainfall Slope stability analyses were conducted for each of the simulations used for seepage analysis, using pore water pressures determined at different time steps as input data for a limit equilibrium analysis performed with SLOPE/W

The X-ray diffraction study confirmed that the soil consists of quartz, feldspar, and illite as major constituent minerals, along with significant amounts of smectite, ka- olinite, calcite, and iron minerals. Large amounts of quartz in the soil imply that the soil is in a later stage of weath- ering after its deposition in the geological past. The presence of illites in the slope materials was the main cause of low frictional resistance obtained by ring shear test. Smectite was also noticed in the soil from X-ray analysis. When such swelling minerals are present in slope materi- als, the slopes are prone to failure. Swelling minerals such as smectites expand when they become wet as water enters the crystal structure and increases the volume of the min- eral. During heavy rainfall, they swell and slopes become prone to failure. Previous studies involving the role of swelling clay minerals in activating a landslide have shown that even a small amount of swelling clay minerals in soil greatly affects its strength behavior. Kerr (1972) high- lighted the role of clay minerals in landslide formation and described the effect on velocity in landslide movement.

Fig 16 Variations of strength of landslide soil with increasing amount of swelling clay minerals (after Yatabe et al. 2000) The bedrock geology of the study area is mainly frac- tured limestone, and on the base of hills villagers use spring water for daily use. Solution activities in limestone facilitate remarkable ground water percolation within the bedrock. Hydraulic conductivity of the rock estimated in this paper by the method proposed by Goodman (1989) suggested saturated flow through the rock mass is possible with a conductivity of 10~* m/s. The orientation of joints in the slope also supports water flow through rock during extreme rainfall events, and such flow was observed during first field visit. The colluvial soil above the bedrock mainly consists of silty clay with fine sand. In situ measurements of infiltration capacity of soils revealed that the soils are relatively permeable and have conductivity rates in the range of 107° m/s. The colluvium is well vegetated by shallow rooted vegetation, is relatively loose, and its bulk density is low. Thus, rainwater can easily infiltrate within the soil. In the extreme rainfall events of 2002, rainfall recharged the fractures in the bedrock and immediately a

Table 1 Result of X-ray analysis of soils collected from both ridges and zero-order basins (2004) in granitic terrains. Wilson and Dietrich (1987) descri bed the bedrock hydrology and pore pressure devel- opment in topographic hollows covered by colluvium. There consid study rock ered during the parameter study area. Hence, the hydraulic cond was estimated according to descri estimate hydraulic conductivity of roc fore, the groundwater flow from fractures needs to be of landslides in the uctivity of the bed- theoretical method bed by Goodman (1989). Equation 4 was used to K:

Cohesion and friction angle were estimated by direct shear test, volumetric water content in saturation and unit weight were estimated in the lab Soil hydraulic conductivity was measured in field. Hydraulic conductivity of rock was estimated from Goodman (1989) Table 2 Parameters used in simulations The volumetric water content functions and the hydraulic conductivity functions were obtained from curves derived using the algorithm of Green and Corey (1971) for soils with similar grain size distributions, adjusting the saturated water content and permeability values to the actual mea- sured values. The main failure events occurred on 23rd July 2002, and it was difficult to describe the initial pressure distribution prior to the rainfall events because almost all Regarding the boundary conditions, a transient flux function with values equal to the daily rainfall rate for July 2002 was applied to the nodes along the ground surface. A null flux condition was assigned to the upslope and sured values. The main failure events occurred on 23rd July Each zero-order configuration was modeled adopting slope angles measured in the field and slope lengths of 20 m. During modeling, all slopes were considered to be geo- metrically planar, having soil layers parallel to the ground surface. The parameters used in the models are illustrated in Table 2.

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Numerical simulation of liquefaction susceptibility of soil interacting by single pile

by Ahmad Asaadi

Previous case histories have shown that soil liquefaction severely damaged many structures supported on pile foundations during earthquakes. As a result, evaluating the potential for instability is an important consideration for the safe... more

Fig. 1. Schematic illustration of FDM model of soil interacting pile in FLAC” Asaadi and Sharifipour / Int. J. Min. & Geo-Eng., Vol.49, No.1, June 2015

Fig.2. Acceleration history and their fast Fourier transforms for: (a) Kocaeli, (b) Kobe, and (c) Bam earthquake 2.3. Analysis approach used as initial stress for the next calculations. Then, the single concrete pile was placed in the soil and the interface properties were applied and thus, the model was analyzed again. In the static analysis, the soil- pile Generally, the model analysis was performed in three phases. First, the soil elements were loaded under a geostatic condition to obtain the natural steady state. These values were

Fig. 4. Variation of maximum R, versus depth in free-field and near pile for (a) Dr=35% , (b) Dr =55% , and (c) Dr =75% during Bam earthquake (PGA =0.29).

The horizontal displacements at the pile head and the settlement time histories of the soil around the pile are presented in Figure 6 (a) and (b), respectively. As expected, by increasing the maximum amplitude of the earthquake, the deformation increased both horizontally and vertically. It was observed that the maximum horizontal displacement at the pile head were 34 and 44 mm for PGA =0.2 and 0.4 g, respectively (absolute peak values). The settlement of the soil around the pile after shaking were 13 mm (for PGA =0.2 g) and 15 mm (for PGA =0.4 g), that is, sizably less than those values for free-field, 80 and 100 mm, respectively (not shown herein). It can be understood that the pile prevents soil settlement because of its usual frictional behavior. Fig. 5. Computed time histories of: (a) excess pore pressure ratio near the pile, and (b) excess pore pressure ratio at free-field

Fig. 6. Computed time histories of: (a) horizontal displacement of pile head, and (b) soil settlement around the pile for different earthquake PGA

Fig. 8. Computed time histories of: (a) horizontal displacement of pile head, and (b) soil settlement around the pile for different earthquake with different predominant frequency

Table 1. Material properties of Nevada sand [16]

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Relationships among remotely sensed soil moisture, precipitation and landslide events

by Ram Ray

2007, Natural Hazards

Landslides are triggered by earthquakes, volcanoes, floods, and heavy continuous rainfall. For most types of slope failure, soil moisture plays a critical role because increased pore water pressure reduces the soil strength and increases... more

Fig. 1 Locations of the three study regions The Cleveland Corral landslide study region in Highway 50 corridor is located in the Sierra Mountains, California, USA (Reid et al. 2003). Highway 50 is a major road located between Sacramento and South Lake Tahoe in California (Spittler and Wagner 1998). The study area is located between 120°17’42” W to 120° 32’42” W and 38° 3912” N to 38° 54’12” N. Altitudes range from about 30-3140 m. Since

Fig. 2 A comparison of soil moisture and rainfall in the three study regions (January 2005- May 2006). Daily values are indicated with circles, triangles, and squares for California, Nepal, and Philippines, respectively. Solid lines correspond to the 1-week moving average

Fig. 3 Slope movement/landslides, 1-week moving average AMSR-E soil moisture and daily TRMM rainfall in California, US from January 2005 to May 2006

Fig. 4 Landslides, lweek-moving average AMSR-E soil moisture and daily TRMM rainfall in Leyte, Philippines from January 2005 to May 2006

Fig. 5 Landslide, 1-week moving average AMSR-E soil moisture and daily TRMM rainfall in Dhading, Nepal from January 2005 to May 2006

Table 1 Statistical analysis of AMSR-E soil moisture from January 2005 to May 2006 in three study regions

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Subsurface deformation measurements during a fast shallow landslide triggered by rainfall

by Devrim Akca

2010, EGU General Assembly (EGU 2010), Vienna, Austria, May 2-7, Vol. 12, p. 15104.

A forested area in Ruedlingen, northern Switzerland, was selected to investigate the geotechnical and hydrological response of a steep slope prior to a rainfall induced failure. Artificial rainfall was applied according to a preplanned... more

A forested area in Ruedlingen, northern Switzerland, was selected to investigate the geotechnical and hydrological
response of a steep slope prior to a rainfall induced failure. Artificial rainfall was applied according to a preplanned
schedule and parameters such as pore water pressure, volumetric water content, horizontal soil pressure,
temperature, piezometric water level and subsurface deformations were monitored. The latter were determined
from four deformation probes that were developed in the Institute for Geotechnical Engineering, ETHZ. Strain
gauges have been attached at a regular spacing along a long, slender, flexible plate to enable measurements of
bending strain to be made at different points along it. The strain gauges were connected as “half bridges” to
minimize the temperature effects. A biaxial inclinometer was also installed on the top of the plate, 20 cm above the
soil surface, to measure the tilt above ground level, providing more boundary conditions to determine the deformed
shape of the probe. The probe is installed vertically inside the soil, while the lowest part is grouted into the stiffer
layer under the topsoil, and is assumed to be stable and without any rotation. Bending strains and the inclination
at the top of the probe are sampled at a frequency of 100 Hz. These are input into an algorithm to determine a
polynomial relationship of deformations and rotations with depth, so that the initiation of slow movements and
propagation of failure during fast soil mass movements can be examined.
A 4-camera arrangement was used for the image acquisition to monitor surface movements using photogrammetric
analyses. Approximately 250 white ping-pong balls were attached to the ground and used as target points. Using a
network simulation tool that was developed in-house, an a priori point positioning accuracy of the ping-pong balls
was estimated to be ± 10.3 mm along the horizontal direction and ± 3.5 mm in the vertical direction. The cameras
operated at a data acquisition rate of circa 8 frames per second (fps). Image measurements were made using the
Least Squares image matching method, which was implemented in another in-house developed software package
(BAAP) to compute 3D coordinates of the balls.
Two sprinkling experiments were conducted in Ruedlingen, in autumn 2008 and spring 2009, the second of which
resulted in mobilising about 130 m3 of debris. In the second sprinkling experiment, the area of interest was moved
ca. 5 metres up the slope. In order to make the targets more discernable on the image space, the ping-pong balls
were replaced with approximately 80 white tennis balls. A posteriori point positioning accuracy obtained from
bundle adjustment in the first sprinkling experiment was ± 16.5 mm along the horizontal direction and ± 3.4
mm along the vertical direction. For the second experiment, these values were ± 11.0 mm and ± 4.3 mm for the
horizontal and vertical directions, respectively.
The results of subsurface deformation during this shallow landslide event are presented and compared with surface
movements determined from photogrammetric measurements.

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Numerical modelling of dam failure due to flow overtopping

by Tawatchai Tingsanchali

2001, Hydrological Sciences Journal

A one-dimensional numerical model for dam failure due to flow overtopping is developed. The MacCormack explicit finite difference scheme is used to solve the one-dimensional equations of continuity and momentum for unsteady varied flow... more

in which the vectors U, F, and S are given by the following equations: The one-dimensional system of equations referred to as steep-slope shallow water equations can be applied for the case of flow over a dam. In vector form, the governing equations are:

Fig. 2 (a) Arc failure surface with flow overtopping and (b) forces considered in ordinary method of slices.

Fig. 3 Flow chart of the numerical computation.

Table 1 Summary of the parameters used in the experiments.

Fig. 5 Model calibration for dam erosion (a) measured and computed reservoir water surtace levels and overtopping discharges and (b) measured and computed dam surface levels. Model verification In verifying the numerical model, the data from experiment al Run B-2 are used. In this experiment, the initial reservoir water level was kept to 5 cm above the dam crest by a vertical plate before allowing overflow to occur. The reservoir inflow was 1.58 | s'. All parameters of the numerical model were kept the same as in the model calibration. From the comparison between the measured and

Considering Model Reach II, the effect of the distance-weighting factor in the numerical scheme in equation (11) was determined first by putting a = 0, 0.003, 0.005, and 0.010. Figure 6(a) shows the effect of @ on the computed dam surface profiles. It was found that @ has an influence on the results especially on the numerical stability and accuracy of the results. When & = 0, the model was unstable in some cases. When a > O, the model was stable and erosion increased with a. When © was increased, e.g. @ = 0.010, the computed bed changes also increased. Sensitivity analysis The experimental data of Run B-1 were used in the study. In order to analyse the effect of the Manning roughness coefficient on the computed results, the values of the roughness coefficient n = 0.010, 0.018, and 0.025 were used in the computation. When Manning n = 0.010 was used (originally = 0.018), it was found that the water depth slightly decreased and the flow velocity increased. Hence, more erosion occurred, especially at the downstream region of dam crest in which the flow velocity is high due to steep slopes. The opposite condition occurs when Manning n is increased, e.g. n = 0.025. However, the effect of Manning n on the computed change in bed level was found to be small.

Model calibration During an overtopping event, two types of fallure modes were observed along the downstream dam slope, i.e. erosion mode and the slope sliding stability mode. The data of Experiments C-2 and C-4 were used in calibrating the model. In the computation of slope stability of the dam, the parameter C,, (equation (22)), which represents the pore water pressure at the base of the failure slice was adjusted. By trial and error, the best agreement between the computed and measured reservoir water levels, overtopping discharges, and dam surface profiles was obtained when C, was 0.80. The other parameters such as the numerical weighting factor a were as same as those found in the dam erosion model calibration. For the case C-2, Fig. 7(a) shows the comparison between the measured and computed reservoir water

SOON agus taen cele emwNeramrers | DMmraeatcy ctr Sag Fig. 9 Effect of slope sliding failure and pore water pressure coefficient on computed dam surface profiles.

* Based on initial upstream water level above dam crest. Fig. 4 Experimental setup.

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Instrumentation for slope stability - Experience from an urban area

by Robin Chowdhury and

1999, Field Instrumentation for Soil and Rock

This paper describes the monitoring of several existing landslides in an urban area near Wollongong in the state of New South Wales, Australia. A brief overview of topography and geology is given and reference is made to the types of... more

Figure 2 - Summary Borehole Log and Inclinometer Profiles for Borehole 4, Site 26 at Coalcliff, near Wollongong in New South Wales, Australia

Figure 3 - Site 026, daily and antecedent rainfall and shear displacement for 3 boreholes. Cumulative displacement and rate of shear for three boreholes at this site, including borehole 4, are shown in Figure 3 together with the 90 day antecedent rainfall curve. The cumulative displacement curves for the three boreholes show the continual movement at the site over the period of monitoring, 28th March 1993 to 9th December 1996. Periods where movement is lower than 0.015 mm per day (the average rate of movement is approximately 0.02 mm per day) correspond to periods of low antecedent rainfall, except Borehole 2 for the period November 95 to February 1996. Movement has continued at this site for the duration of the whole monitoring period.

Figure 5 - Summary borehole log and inclinometer profiles for borehole 13, Site 134 at Woonona Heights, near Wollongong in New South Wales, Australia. The Wonoona Heights Landslide has a plan area of approximately 20,500 m?, a volume of approximately 225,000 m° and is ranked in the land instability database as the 9th largest landslide within the subject area. The landslide site is located within a densely developed urban area, which contains up to 100 residential houses, 29 of which are situated within the landslide. Another 10 houses are located immediately adjacent to the margins of the landslide. One house has been destroyed by ground movement, whilst at least 19 have required major repairs.

Figure 1 - Study Area Location Plan showing the extent of Triassic and Permian sediments in the Southern half of the Sydney Basin with a low angle dip, generally less than five degrees, towards the northwest. This gentle northwesterly dip is a result of the relative position of the district on the southeastern flanks of the Sydney Basin. The northwesterly dip is superimposed with relatively minor syn-depositional and post-depositional structuring (folding and faulting). Normal faulting within the Illawarra area is common, although the fault throws infrequently exceed 5 metres. The structural geology of Wollongong, Kiama and Robertson, mapped on 1:50000 sheets, has been discussed in detail in Bowman (1974).

Table 1 - A Sample of 10 Boreholes out of a total of 42 in the borehole inclinometer dataset with site reference code (SRC), monitoring coverage and rates of movement. It was considered necessary to adopt a standard numbering system for the 328 sites of instability incorporated into a landslide inventory of this area (Chowdhury and Flentje, 1997, 1998a). This landslide inventory was compiled during the research project mentioned above. Each site is identified with a 3 character Site Reference Code (SRC). This site reference code is unique to each location, is plotted for each site on a map record and constitutes the “primary” or “key” field in a land instability database.

Table 2 - Borehole 4, Site 026, monitoring period and displacement (A axis data only) recorded at 4.5m depth Total displacements indicated for boreholes 2, 4 and 6 over the period 28th March 1993 to 9th December 1996, are 27.6 mm, 29.4 mm and 29.0 mm respectively, all at average rates of approximately 0.02 mm per day, or, if extrapolated, 8 mm per year. On the WP/WLI (1993) velocity scale, such a rate of displacement is classified as extremely slow. Y et, RSA engineers maintaining the dual electric railway line that traverses this site have reported an annual maintenance cost exceeding tens of thousands of dollars.

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The South Pyrenean Eocene carbonate megabreccias revisited: new interpretation based on evidence from the Pamplona Basin

by Aitor Payros

1999, Sedimentary Geology

The South Pyrenean Foreland Basin contains numerous units of Eocene carbonate megabreccias intercalated with siliciclastic turbidites and derived by resedimentation of shallow-marine carbonate platforms. Previous studies were limited... more

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Relationships between tectonics, slope instability and climate change: Cosmic ray exposure dating of active faults, landslides and glacial surfaces in the SW Alps

by michel corsini

2010, Geomorphology

In the Argentera massif (French Southern Alps), large active landslides develop along strike of an active corridor of dextral strike-slip faults revealed by shallow ongoing seismicity. Glacially polished bedrock outcrops are offset by... more

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Loading frequency effect on stiffness, damping and cyclic strength of modeled rockfill materials

by Abbas Ghalandarzadeh

2012, Soil Dynamics and Earthquake Engineering

This paper studies the loading frequency and waveform effects on shear modulus (G) and damping ratio (D) of three high compacted modeled rockfill materials by conducting large-scale triaxial testing. The laboratory test results have shown... more

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EFFECT OF NONLINEARITY ON DYNAMIC RESPONSE OF EARTHEN DAM

by Mohammad Asif

The stability and safety are very important issues for the dam structure which are built in seismic regions. The dam body consist of soil materials that behave nonlinearly need to be modelled with finite elements. In present study, the... more

To compute stresses correctly, it is necessary to apply the weight of the reservoir as boundary condition. The soil in the core and shell region are assigned by the linear and elastic plastic soil model to observe effect of nonlinearity. Formulation In finite element treated as two d analysis of dam, the problem is imensional and the dam cross- section was represented as assemblage of constant strain triangular element used in t and quadrilateral elements. The he present investigation is a plain strain quadrilateral element composed of two four nodal point triang es.

n Eq. 1, [M], [C] and [K] are the mass matrix, viscous damping matrix and stiffness matrix, respectively for the finite element system. {r} is the vector of 2N nodal points in the finite element idealization: {r} and { f } are respectively the nodal point velocity and acceleration vectors. In the case of earthquake excitation the load vector {R(t)} is a function of the nodal point masses and accelerations.

Figure 5: The contours of vertical stresses of (a) linear and (b) nonlinear soil model.

Figure 6: The contours of horizontal stresses of (a) linear and (b) nonlinear soil model

Table 2: Results of the Nara Dam for Seismic Analysis CONCLUSIONS nonlinear soil mod el it is while minimum vertical s found to be at the and -8.32 kPa, res It is observed that in both the cases, the horizontal stresses increase from top to bottom of the dam which is as expected. The maximum horizontal stress for linear soil mod el is 753.69 kPa and for 808.94 kPa. This indicate that as the nonlinearity increases, stresses increases tresses for these cases are toe of dam which are -14.35 kPa pective y. It shows that effect of tension decreases with nonlinearity of the soil. The effect of nonlinearity was investigated on vertical as well as horizontal displacements. It was observed that both displacements increases due to nonlinearity. All the results of the seismic analysis of Nara Dam is summarized in Table: 2 shown below. It can be observed that the value of stresses and displacement increase due to the effect of nonlinearity of the soil.

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Finite element formulation and algorithms for unsaturated soils. Part II: Verification and application

by Antonio Gens

2003, International Journal for Numerical and Analytical Methods in Geomechanics

This paper presents a complete finite-element treatment for unsaturated soil problems. A new formulation of general constitutive equations for unsaturated soils is first presented. In the incremental stress-strain equations, the suction... more

Figure 1. Yield surface intersection: elastic to plastic transition. has changed from elastic to plastic. Such a change occurs if

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Nature and timing of large landslides in the Himalaya and Transhimalaya of northern India

by Lewis A Owen

2008

Four large landslides, each with a debris volume >10 6 m 3 , in the Himalaya and Transhimalaya of northern India were examined, mapped, and dated using 10 Be terrestrial cosmogenic radionuclide surface exposure dating. The landslides date... more

Four large landslides, each with a debris volume >10 6 m 3 , in the Himalaya and Transhimalaya of northern India were examined, mapped, and dated using 10 Be terrestrial cosmogenic radionuclide surface exposure dating. The landslides date to 7.7AE1.0 ka (Darcha), 7.9AE0.8 ka (Patseo), 6.6AE0.4 ka (Kelang Serai), and 8.5AE0.5 ka (Chilam). Comparison of slip surface dips and physically reasonable angles of internal friction suggests that the landslides may have been triggered by increased pore water pressure, seismic shaking, or a combination of these two processes. However, the steepness of discontinuities in the Darcha rock-slope, suggests that it was more likely to have started as a consequence of gravitationally-induced buckling of planar slabs. Deglaciation of the region occurred more than 2000 years before the Darcha, Patseo, and Kelang Serai landslides; it is unlikely that glacial debuttressing was responsible for triggering the landslides. The four landslides, their causes, potential triggers and mechanisms, and their ages are compared to 12 previously dated large landslides in the region. Fourteen of the 16 dated landslides occurred during periods of intensified monsoons. Seismic shaking, however, cannot be ruled out as a mechanism for landslide initiation, because the Himalaya has experienced great earthquakes on centennial to millennial timescales. The average Holocene landscape lowering due to large landslides for the Lahul region, which contains the Darcha, Patseo, and Kelang Serai landslides, is w0.12 mm/yr. Previously published large-landslide landscape-lowering rates for the Himalaya differ significantly. Furthermore, regional glacial and fluvial denudation rates for the Himalaya are more than an order of magnitude greater. This difference highlights the lack of large-landslide data, lack of chronology, problems associated with single catchment/large landslide-based calculations, and the need for regional landscape-lowering determinations over a standardized time period.

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In Situ Pore Water Pressure Dissipation Testing of Marine Clay under Reclamation Fills

by Myint Bo

2006, Geotechnical and Geological Engineering

In situ dissipation tests provide a means of evaluating the in situ coefficient of horizontal consolidation and horizontal hydraulic conductivity of soft clays. Dissipation tests by means of piezocone (CPTU), dilatometer (DMT),... more

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Analysis of debris flow behavior with a one dimensional run-out model incorporating entrainment

by Jean-philippe Malet and

2012, Engineering Geology

a b s t r a c t a r t i c l e i n f o Available online xxxx Keywords: Debris flows Run-out Entrainment Erosion Modeling

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Quantifying the mechanical and hydrologic effects of riparian vegetation on streambank stability

by Andrew Collison

2002, Earth Surface Processes and Landforms

Riparian vegetation strips are widely used by river managers to increase streambank stability, among other purposes. However, though the effects of vegetation on bank stability are widely discussed they are rarely quantified, and... more

Riparian vegetation strips are widely used by river managers to increase streambank stability, among other purposes. However, though the effects of vegetation on bank stability are widely discussed they are rarely quantified, and generally underemphasize the importance of hydrologic processes, some of which may be detrimental. This paper presents results from an experiment in which the hydrologic and mechanical effects of four riparian tree species and two erosion-control grasses were quantified in relation to bank stability. Geotechnical and pore-water pressure data from streambank plots under three riparian covers (mature trees, clump grasses and bare/cropped turf grass) were used to drive the ARS bank stability model, and the resulting factor of safety (F s ) was broken down into its constituent parts to assess the contribution (beneficial or detrimental) of individual hydrologic and mechanical effects (soil moisture modification, root reinforcement and surcharge). Tree roots were found to increase soil strength by 2-8 kPa depending on species, while grass roots contributed 6-18 kPa. Slope stability analysis based on data collected during bank failures in spring 2000 (following a very dry antecedent period) shows that the mechanical effects of the tree cover increased F s by 32 per cent, while the hydrologic effects increased F s by 71 per cent. For grasses the figures were 70 per cent for mechanical effects and a reduction of F s by 10 per cent for the hydrologic effects. However, analysis based on bank failures in spring 2001 (following a wetter than average antecedent period) showed the mechanical effects of the tree cover to increase F s by 46 per cent, while hydrologic effects added 29 per cent. For grasses the figures were 49 per cent and 15 per cent respectively. During several periods in spring 2001 the hydrologic effects of the tree cover reduced bank stability, though this was always offset by the stabilizing mechanical effects. The results demonstrate the importance of hydrologic processes in controlling streambank stability, and highlight the need to select riparian vegetation based on hydrologic as well as mechanical and ecological criteria. Published in

Figure 2. Relationship between root diameter and tensile strength for six vegetation species

Figure 3. Distribution of root numbers with depth, subdivided by root diameter

Figure 4. Distribution of root area ratio with depth, subdivided by root diameter

Figure 5. Distribution of cohesion due to roots with depth, subdivided by root diameter

Figure 6. Pore-water pressure response to rainfall for the three streambank vegetation covers Comparison of the tensiometer results from 100 cm reveals significant differences between all three covers. The control site and the grass plot broadly track each other, with the control consistently maintaining a suction around 10 to 20 kPa higher than the grass, indicating that stemflow may be concentrating water to depth in the grass plot (Figure 6, top right). This is even more pronounced in the tree plot, further pointing to the potential negative effects of rapid delivery of water along root- and faunal-induced macropores. Evidence for preferential macropore flow is not confined to the vegetation covers, however, with all three treatments showing response times to the 4 April event that exceed matric conductivities.

Figure 7. Factor of safety versus time for the three streambank vegetation covers

Figure 8. Minimum factor of safety for all species during a dry year (spring 2000), subdivided by vegetation effects

Figure 9. Minimum factor of safety for all species during a wet year (spring 2001), subdivided by vegetation effects

Figure 10. Factor of safety for the mixed tree cover and bare plot (December 1999—March 2001), subdivided by vegetation effects

Table I. Soil properties used in bank stability model Increases in cohesion due to roots are added to Lower Holocene (a) scaled to allow for thickness. Lower Holocene (b/c) is divided to allow for different pore-water pressures.

Earth Surf. Process. Landforms 27, 527-546 (2002) Table II. Summary of root properties * Root diameter (mm). ¥ Root tensile strength (MPa). Roo, Depth above which 90 percent of roots are found (cm). At/A, Area of roots per unit area, averaged over 1 m depth of soil. c,, Additional cohesion due to roots (kPa) averaged over 1 m depth of soil.

Table III. Comparison of vegetative effects on bank factor of safety (F,) during critical conditions (bank failure) for a relatively dry year (April 2000) and year (February 2001) Based on average root reinforcement values for sweetgum, river birch and sycamor Based on matric suction recorded under riparian trees. Based on matric suction recorded under gamma grass. Earth Surf. Process. Landforms 27, 527-546 (2002)

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Monitoring and analysis of settlement and stability of an embankment dam constructed in stages on soft ground

by Azm S Al-Homoud

2001, Bulletin of Engineering Geology and the Environment

Fig. 1 used; (2) drained (long-term) where effective stress anal- ysis (ESA) is used; and (3) partially drained (intermediate) using a combination of (1) and (2), ie. the undrained strength analysis (USA). Details of these methods are given

Fig. 3 Project layout (Sir Alexander Gibb and Partners 1986a)

an be Main zones of a typical cross section of the Karameh dam project. 973). (Modified after Sir Alexander Gibb and Partners 1996) tion that occurs during construction (Ladd 1991). Assuming that significant consolidation does not occur as the fill is placed, the excess pore-water pressure developed in the foundation soils beneath the embankment can be estimated using the methods developed by Lambe (1973).

Typical cross section in dam body and foundation at Ch 1480 Fig. 6

Fig. 5 a pe Plan of the Karameh dam project showing types of earthfill mate- rial used in construction (Sir Alexander Gibb and Partners 1993)

Case study: Karameh embankment dam Fig. 7 - Cross section showing settlement cells in foundations and piezometers in fill and foundations at Ch 1480 in the Karameh dam. (Modified After Sir Alexander Gibb and Partners 1997)

settlement for a longitudinal section of the Karameh dam using different models. characteristics of the A dish shape is seen, due to the foundation and the location of maximum loading along the dam axis - the location of maximum fill is approximately at the centre of the dam. Due to the topography o is shorter drainage path drainage and dissipatio ikely that the geologica f the site, towards both ends, there under the crest resulting in lateral n of excess pore pressure. It is conditions may have an impor- tant role in creating this dish shape as many valleys have more compressible material in the valley floor. Figure 20 gives the predicted and measured consolidation Variation in piezometric readings with time at Ch 1480 (Sir Alex- ander Gibb and Partners 1997)

Limiting equilibrium stability analysis is used to obtain a quantitative measure of safety using field instrumentation data. The effective stress analysis (ESA) could be used to check stability during construction based on measured pore pressures, whereas a total stress analysis (TSA) can be used based on subsequent field vane testing of undisturbed samples. prey 2 a Toe eB 1 é xo: — om: x os i.

Fig. 11 used. The effective shear strength par ameters used in the analysis were based on the results shown in Tables 3-14. The initial pore-water pressure param be 0.7 in the core. Typical results ical failure surface shown in Fig. 22. 7 upper bound of computed safety fa using ESA and field data pore-water foundation. The lower bound of com eter r, was taken to of the analysis are presented for the dam section at Ch 1480. Figure 22 shows the critical failure surface for USA and ESA, while Fig. 23 provides the safety factor versus height of fill for the crit- The evaluated safety factors tend to decrease with increasing height of fill. The ctors was obtained pressures, while the ower bound was computed using ESA with B=0.9 in the puted safety factors seems to be reasonable. These results indicate that the dam should be stable during construction with a controlled rate of loading. Fill elevation and settlement readings from settlement cells in foundation and fill at Ch 1480 (Sir Alexander Gibb and Partners 1997)

Detailed cross section of dam body at Ch 1480

Variation in effective overburden stress, preconsolidation pres- sure and effective final stress under foundation at Ch 1480 at different stages of fill foundation material. Typical safety factor values for critical failure surface versus height of fill are shown in Fig. 23. The results were obtained based on the monitored loading- induced pore-water pressure. The evaluated safety factors were compared with those suggested by Sir Alexander Gibb and Partners (1997) for the final stage of construction Variation in undrained shear strength (S,,) of foundation at Ch 1480 with depth at different stages of fill

Fig. 15 Fig. 17 Predicted and measured settlement of foundation vs. effective stress at Ch 1480 (total depth of stressed sub-layers under foun- dation = B/2, B=total width of loaded area) Exponent parameter (m) and normally consolidated strength value (S) for normalized undrained strength relations of founda- tion material coefficient of compressibility) such that when fill place- ment is resumed, the induced pore water pressures are lower than those that would otherwise have developed. The rate of consolidation of partially saturated soil is relatively

0 m Variation in measured and predicted settlement under founda- tion with depth at Ch 1480 (total depth of stressed sub-layers under foundation =B/2, B =total width of loaded area)

Instantaneous loading, predicted and measured settlement of foundation vs. time at Ch 1480 (total depth of stressed sub-layers under foundation =B/2, B =half total width of loaded area)

A detailed geotechnical analysis was carried out of the monitoring data obtained for the Karameh embankment dam which was constructed in stages on soft ground. Controlled rates of loading were determined in order to attain sufficient soil strengthening such that the maximum required load could be safely supported. The geological and geotechnical information obtained included an engi- neering geological assessment of the site, laboratory and in-situ tests (consolidation, shear strength, pressure development) as well as physical and pore water index prop- erties. During the construction of the dam, settlement and Predicted and measured total overall settlement (initial settle- ment, consolidation of foundation and settlement of fill) vs. chai- nage using different models (total depth of stressed sub-layers under foundation = B/2, B =total width of loaded area)

foundation were monitored. The overall stability of the Karameh dam embankment and foundation was evaluated using ESA and TSA methods. The monitored data were compared to those obtained using soil mechanics prediction models. Three main sections along the Karameh dam were analysed in detail:

Instantaneous loading, predicted and measured settlement of fill at Ch 1480 using the m, method

plastic limit (%); PI plasticity index (%); % <0.074 percent passing 0.047-mm sieve; % <0.002 percent passing 0.002-mm sieve; SG specific gravity Summary of classification test results for foundation materials (Sir Alexander Gibb and Partners 1986c). SD Standard deviation; NMC natural moisture content (%); LL liquid limit (%); PL

Permeability values for foundation and embankment material. (Sir Alexander Gibb and Partners 1988)

* Mixed sequences of clays, silts, sands and gravel with laminated soil, with a thickness of >80 m. Geologists suggest that a depth to the Ghor El Katar series of 200-300 m may be appropriate. Thick- ness of Quaternary deposits is 250 m; therefore that of Samra Formation is 200 m Thickness and elevation (both in m) of main units (Sir Alexander Gibb and Partners 1993)

Material densities of foundation material (Sir Alexander Gibb and Partners 1988) Preconsolidation pressure (o;) for different foundation strata at the main studied sections based on oedometer tests (Sir Alex- ander Gibb and Partners 1993). OGL Original ground level

* Lower bound derived from consolidation undrained (CU) test > Upper bound Shear strength test results and density for fill materials. (Sir Alexander Gibb and Partner 1993)

* Upper bound > Lower bound Shear strength test results for foundation materials (Sir Alexander Gibb and Partners 1988 Table 4

Average design values for coefficient of consolidation (c,) and coefficient of compressibility (m,) for foundation material and wet core. (Sir Alexander Gibb and Partners 1993)

* At OMC+4% > At NMC (natural moisture content) + 4% Consolidation and compressibility parameters for compacted fill material (Sir Alexander Gibb and Partners 1988). OMC Optimum moisture content

* Effective values Summary of principal design parameters of embankment mate- rial (Sir Alexander Gibb and Partners 1986b)

Piezometers in the Karameh dam project Table 15 models. Control samples were used to evaluate the accu- racy of the monitored data and calibrate the prediction models including parametric studies. Based on the outcomes of this comprehensive exercise, the stability of the dam was evaluated during construction. angle of friction, compressibility and consolidation coeffi- cients, etc., together with the pore-pressure values prior to loading during the staged construction of the dam. The upper bound, lower bound and design values of undrained shear strength for the Lisan foundation material are given in Table 13 and the main consolidation parameters for the core and foundation of the dam in Table 14.

Piezometric readings, fill levels, total vertical stress change, B or r, values, effective stress increase, current effective stress and estimated preconsolidation pressure for meters at Ch 1480. (Sir Alexander Gibb and Partners 1997)

Predicted consolidation settlement of foundation at Ch 1480 using Terzaghi (1943) consolidation theory, 2-D and 3-D consolidati theories for three stages of loading. EOC End of construction; p;,, total consolidation settlement

* Terzaghi (1943) theory > Skempton and Bjerrum (1957) theory Summary of predicted total overall settlement of foundation (initial and consolidation settlement of foundation and fill settlement) at different stations using different models. EOC End of construction Table 19

Predicted and measured consolidation settlement of foundation at different offsets for cross section at Ch 1480. All measurements are in metres. EOC End of construction; DFSS double fluid settlement system Table 20

* For anisotrophic permeability > For isotropic permeability Results of parametric analysis of effect of lateral drainage on horizontal coefficient of consolidation (C,), and ratio of hori- zontal to vertical permeability (K,/K,) at Ch 800, 1480 and 1850. K,/K, Ratio of horizontal to vertical permeability; C, horizontal coefficient of consolidation; U% percent consolidation; H, effec- tive drainage thickness; H, thickness of drainage layer; ts. time at 50% consolidation

Back analysis of monitored data of settlement and pore water pressure for evaluating field values of C,, m, and c,, p, Consolidation settlement; H, thickness of drainage layer Table 21

Main input parameters of undrained shear strength (S,) at different stages of fill for selected sections of Karameh dam Table 23

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The 17 February 2006 Guinsaugon rock slide-debris avalanche, Southern Leyte, Philippines: deposit characteristics and failure mechanism

by Mark Zarco

2008, Bulletin of Engineering Geology and The Environment

The Guinsaugon rock slide-debris avalanche was the most catastrophic single landslide event in Philippine history, with 14–18 M m3 of debris instantly burying an entire village. Hummocky topography, pressure ridges and other internal... more

The Guinsaugon rock slide-debris avalanche was the most catastrophic single landslide event in Philippine history, with 14–18 M m3 of debris instantly burying an entire village. Hummocky topography, pressure ridges and other internal structures suggest that the landslide deposit was emplaced as a debris avalanche and debris flow. Susceptibility to planar and wedge failures as well as to toppling due to rock discontinuities were demonstrated using kinematic analysis and SMR. Limit equilibrium analysis on planar failures yielded factors of safety ranging from 0.8 to 1.5. The study showed that pore pressure on discontinuities had a more significant effect on the slope stability than seismicity. For wedge failures, there is a sudden drop in the factor of safety at pore water pressures of 258–306 kPa. At the site, the pore water pressure may have been as high as 490 kPa. The possibility of such a landslide event in the future is discussed. Le glissement rocheux/avalanche de roches de Guinsaugon est le glissement le plus catastrophique de l’histoire des Philippines, avec 14 à 18 M m3 de débris rocheux submergeant entièrement un village. La topographie chaotique, les rides de pression et différentes structures internes suggèrent que les dépôts se sont mis en place comme une avalanche de roches ou un écoulement de débris. La prédisposition du massif rocheux aux glissements plans ou en dièdres, ainsi qu’aux basculements a été établie à partir de l’analyse des discontinuités et en considérant la classification SMR. L’analyse suivant la méthode des équilibres limites a conduit à des facteurs de sécurité variables entre 0,8 et 1,5 pour les glissements plans. Les études ont montré que les pressions interstitielles dans les discontinuités avaient une plus forte influence sur la stabilité que les actions séismiques considérées. Pour les ruptures en dièdres, une baisse rapide des coefficients de sécurité apparaît lorsque les pressions interstitielles passent de 258 à 306 kPa. Sur site, les pressions interstitielles peuvent avoir atteint la valeur de 490 kPa. La possibilité d’un événement comparable à l’avenir est discutée.

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Multiple edifice-collapse events in the Eastern Mexican Volcanic Belt: The role of sloping substrate and implications for hazard assessment

by Michael F Sheridan

2006, Journal of Volcanology and Geothermal Research

The Citlaltépetl-Cofre de Perote volcanic chain forms an important physiographic barrier that separates the Central Altiplano (2500 masl) from the Gulf Coastal Plain (GCP) (1300 masl). The abrupt eastward drop in relief between these... more

The Citlaltépetl-Cofre de Perote volcanic chain forms an important physiographic barrier that separates the Central Altiplano (2500 masl) from the Gulf Coastal Plain (GCP) (1300 masl). The abrupt eastward drop in relief between these provinces gives rise to unstable conditions and consequent gravitational collapse of large volcanic edifices built at the edge of the Altiplano. Eastward sloping substrate, caused by the irregular configuration of the basement rocks, is the dominant factor that controls the direction of collapsing sectors in all major volcanoes in the region to be preferentially towards the GCP. These collapses produced voluminous debris avalanches and lahars that inundated the well-developed drainages and clastic aprons that characterize the Coastal Plain. Large catastrophic collapses from Citlaltépetl, Las Cumbres, and Cofre de Perote volcanoes are well documented in the geologic record. Some of the avalanches and transformed flows have exceptionally long runouts and reach the Gulf of Mexico traveling more than 120 km from their source. So far, no direct evidence has been found for magmatic activity associated with the initiation of these catastrophic flank-collapses. Apparently, instability of the volcanic edifices has been strongly favored by very intense hydrothermal alteration, abrupt topographic change, and intense fracturing. In addition to the eastward slope of the substrate, the reactivation of pre-volcanic basement structures during the Late Tertiary, and the E-W to ENE-SSW oriented regional stress regimes may have played an important role in the preferential movement direction of the avalanches and flows. In addition to magmatic-hydrothermal processes, high amounts of rainfall in the area is another factor that enhances alteration and eventually weakens the rocks. It is very likely that seismic activity may be the principal triggering mechanism that caused the flank collapse of large volcanic edifices in the Eastern Mexican Volcanic Belt. However, critical pore water pressure from extraordinary amounts of rainfall associated with hurricanes or other meteorological perturbation cannot be ruled out, particularly for smaller volume collapses. There are examples in the area of small seismogenic debris flows that have occurred in historical times, showing that these processes are not uncommon. Assessing the stability conditions of major volcanic edifices that have experienced catastrophic sector collapses is crucial for forecasting future events. This is particularly true for the Eastern Mexican Volcanic Belt, where in many cases no magmatic activity was associated with the collapse. Therefore, edifice failure could occur again without any precursory warning.

Fig. 1. Thematic Mapper satellite image of the eastern Mexican Volcanic Belt (MVB) courtesy of M. Abrams at Jet Propulsion Laboratory, USA. Inset box shows image location in the MVB.

Fig. 2. Photographs of Citlaltépetl volcano. (A) Panoramic air photo of Citlaltépetl showing the present cone and the ancient cone remnants. (B) Vie" of the interior of Citlaltépetl’s crater showing areas of fresh and altered rocks. Cerro Desconocido and Cerro Tecomales, which are of probable Plio-Pleistocene age, form the La Gloria Volcanic Complex. Due to their morphology (see Fig. 1), they have been regarded as eroded calderas (Negendank et al., 1985); however, we have found no evidence to support that hypothesis, so for the present they can be considered as an eroded chain of aligned vents. Citlaltépetl, also known as Pico de Orizaba, is North America’s highest volcano (5685 m) and Mexican’s highest peak. This ice-capped, andesitic stratovolcano, located at the southern end of the CCPVR, is the only historically active volcano in the area. Its most recent

Fig. 3. Simplified geologic map of Citlaltépetl volcano showing the three main eruptive stages and the surrounding volcanic features (outer silicic domes and Sierra Negra volcano) modified from Carrasco-Ntiiez (2000). About 7 km southwest of Citlaltépetl is Sierra Negra Volcano, a 4650-m-high andesitic stratovolcano which first erupted during the Pleistocene and has a possible age of about 0.5 Ma (Hoéskuldsson, 1992). Sierra Negra is not aligned in the main N-S volcanic trend, but has a ocal lineament with Citlaltépetl and some dacitic domes along a NE-SW direction. Some additional dacitic and rhyolitic domes are situated between the Las Cumbres and Citlaltépetl centers with no particular alignment. The geologic record of these volcanoes show that several catastrophic edifice collapses have occurred, and in the cases of Citlaltépetl and Cofre de Perote there have been multiple events. Some of the resulting avalanches and transformed flows have exceptionally long runouts and reach the Gulf of

Fig. 4. Distribution of the main debris-avalanche and debris-flow deposits derived from the Citlaltépetl—Cofre de Perote Volcanic Range.

Fig. 5. Photograph showing a 5-m high hummock of the Jamapa avalanche in the Huatusco area (see Fig. 4 for location’ The Teteltzingo avalanche-lahar appears to be a single massive, unbedded, poorly sorted mixture of heterolithologic pebbles, cobbles, and boulders sup- ported within a characteristic yellow-brown, clayey, silty sand matrix (Fig. 6B) that contains small vesicles suggestive of air bubbles trapped in a water-saturated matrix. The deposit is 12—20 m thick on average, but in a few places is up to 100 m thick. Veneer deposits extend up to 60 m above the terrace level in proximal facies. This clay-rich lahar contains 10-16% of secondary alteration minerals such as smectite, and kaolinite, an assemblage that is comparable with other cohesive lahars such as the Osceola lahar (formerly “mudflow”, Crandell, 1971) that originated from the collapse of This is a small debris avalanche deposit, probably less than 1 km?, which fills the Metlac valley with a thickness of more than 100 m, but has a very limited distribution (Fig. 4). It is an indurated, massive, bouldery-rich, matrix-supported, heterolithologic deposit, which is dominated by andesitic clasts. The deposit has a very irregular and conspicuous shape due to erosion along its margins on both sides of the main river-valley. This morphology may suggest that the

Fig. 6. Photographs of the Teteltzingo avalanche-lahar deposit and its source area. (A) The upper remnants of the Espolon de Oro peak, the source ar for the Teteltzingo lahar showing highly altered layers that alternate with fresh lava flows. (B) A 25-m-thick massive clay-rich deposit near Cordol City, at a distance of about 45 km from source, but still containing very large blocks.

Fig. 7. Plot of the degree of hydrothermal alteration versus the area covered by glacial ice for the Cascade volcanoes and Citlaltépetl (from Carrasco-Niiiez et al., 1993). 0—no alteration; 1—small areas of moderate alteration; 2—moderate alteration; 3—large areas of moderate alteration; 4—large intensely altered rocks. C—Citlaltépetl, GP—Glacier Peak; MA—Mount Adams; MB—Mount Baker; MH— Mount Hood; MJ—Mount Jefferson; ML—Mount Lassen; MR— Mount Rainier; MS—Mount Shasta, MSH—Mount St. Helens; TS— Three Sisters.

Fig. 8. Photograph of Las Cumbres Avalanche deposit in a proximal area, about 12 km from source, showing the chaotic distribution of clasts and highly altered areas. The white bar is 2 m long.

Fig. 9. Photograph showing the 1920 Huitzilapan deposit overlaying the 5860+ yr B.P. debris flow deposit along the Huitzilapan river-valley. Dashed line indicates the contact between these two deposits.

Fig. 10. Comparison of photographs of the 1920 seismogenic debris flow taken in 1920 and 1999, showing some landslide scarps and the lower terrace formed by the 1920 seismogenic debris flow along the Huitzilapan River.

Fig. 11. Photographs of the Xico avalanche and its source area. (A) View of the eastern flank of Cofre de Perote summit area showing the horseshoe- shaped amphitheater. (B) Outcrop at about 10 km from source showing the massive and chaotic nature of the Xico avalanche deposit with large blocks within a silty matrix.

This deposit consists of at least two different units that crop out along Los Pescados river valley forming a central terrace slightly dipping to the east. The deposit forms the lowermost channel-filling terrace deposit of Los Pescados River, and is therefore the youngest deposit in the area (Fig. 12B and cf. Hubbard et al., in press). It overlies a basaltic lava plateau, which has been The deposit consists of massive, heterolithologic mixture of boulders and coarse gravels within a silty— clayey matrix. Some blocks show jigsaw-fractures Fig. 12. Photographs of Los Pescados deposit. (A) Outcrop along Los Pescados River, about 30 km from source, showing a 1.4 m long block witt jigsaw fracturing. (B) Panoramic view of Los Pescados northern river wall at the confluence with the Huitzilapan drainage, showing the Cretaceous limestone wall. The river-valley filled by an upper terrace composed of a horizontal basaltic lava flow that was incised and later filled by the Los Pescados deposit.

Fig. 13. Distribution of the limestone basement rocks (modified from Diaz-Castellon, 2003). In the Serdan-Oriental basin they are partially buried by pyroclastic and lacustrine deposits and have a higher altitude in comparison with equivalent rocks outcropping on the coastal plain. Section a—a' is shown in Fig. 14. Instability can occur rapidly (months or up to a few years) associated with a discrete geologic event such as magmatic intrusion or a large seismic shock as in the case of Mount St. Helens in 1980 (Christiansen and Peterson, 1981) or the eruption of Bezymianny volcano (Belousoy, 1996). However, instability can also develop over long periods of time due to the cumulative effect of several individual destabilizing factors, creating progressively unstable conditions that can be accelerated by the action of a sudden discrete event, such an earthquake. There are several factors that control edifice instability; some are related to magmatic activity like direct magma intrusion (Swanson et al., 1976; Elworth and Voight, 1996) or phreatomagmatic activity (Moriya,

Based on the orientation of cone alignments follow- ing Nakamura (1977), we infer that the regional maximum horizontal stress direction is E-W in the southern portion of the eastern MVB, following the same trend of cone alignments and active faulting directions as in the central MVB reported by Suter et al. (1992). However, in the central and northern parts of the eastern MVB these trends are rotated to the ENE direction (Siebert and Carrasco-Nufiez, 2002). These faulting has been reported. However, some fracturing patterns and lineaments can be distinguished from satellite imagery interpretation (Negendank et al., 1985; Concha-Dimas et al., 2005) as shown in Fig. 15. A dominant NW direction corresponds with the structural domain that affects mostly Mesozoic sedi- mentary rocks. The two other important patterns oriented either E-W or ENE, both affect Quaternary volcanism. The CCPVR has been regarded as a N-S alignment of volcanoes. However, a careful inspection of the principal vent distribution (Fig. 1) clearly shows that the large volcanoes in fact follow two main Fig. 15. Structural map showing the main fracture systems, volcano vent alignments, and distribution of earthquake epicenters (modified from Diaz. Castellon, 2003).

Fig. 16. (A) Alteration mineral map for Citlaltépetl (Pico de Orizaba) volcano using AVIRIS imagery (see text for details). (B) Remote sensing base estimates of halloysite in soil and weathered volcanic rocks of La Gloria area (see text for further details).

Fig. 17. Slope map of the eastern MVB indicating three different ranges of values: 15—30° for intermediate slope, 30—45° for high slope, and 45—60° for very high slopes that exceed all repose angles. In contrast, the Teteltzingo deposit is a clay-rich lahar (from Citlaltépetl volcano) that originated by the rapid transformation of an avalanche descending from the upper summit area of an ancestral cone (Espolén de Oro). Evidence for this conclusion is the extensive areas of intense hydrothermally altered rocks within the deposit. It is very likely that instability developed progressively over long periods of time, but the process could have been accelerated by a sudden event such an earthquake and/or extraordinary periods of heavy rainfall. In fact, landslides triggered by earthquakes are relatively common (Keefer, 1984; Schuster and Cran- dell, 1984). In cases such as the 1994 earthquake that occurred near the Nevado de Huila volcano in Colombia (Scott et al., 2001), thousands of small areas saturated by prolonged rainfall were ready to be mobilized. Although the resulting seismogenic flow along the Paez River had a relatively low volume, it produced a catastrophe because it traveled more than 100 km as it caused as many as 1000 casualties. The casualties estimated during the 1920 Huitzilapa debris flow range

Fig. 18. Hazard map for debris flows of the CCPVR (modified from Diaz-Castellon, 2003) showing three different levels of hazard. High level is indicated in dark pattern, medium level is in grey pattern, and low level is in light colored pattern (see text for more details). Because Citlaltépetl is the only large active volcano of the CCPVR it should be considered as the most hazardous edifice in the region. This is particularly true due to its passive fumarolic activity, its present highly unstable summit cone with very steep slopes, and its large active glacier that can provide water saturation conditions. Its summit area is strongly fractured and shows extensive zones of hydrothermally altered rock masses (e.g., Carrasco-Nufiez, 2000; Hubbard, 2001; masses (e.g., Carrasco-Nufez, 2000; Hubbard, 2001; A slope map can rapidly identify and classify unstable areas (Fig. 17). Trajectories of potential debris flows can be delineating along the main drainage network by modeling different degrees of hazard. The model for this paper was generated by using the Flow- 2D software (unpublished by Sheridan, 1986), which was adapted to work in a MS-Windows environment, using 1:250,000 scale cartography and Digital Elevation

The collapse direction was solely towards the eastern coastal plain, principally in response to the highly Finally, it is important to emphasize that the failure of a volcanic edifice can occur without warning. This is particularly the case where there is clear evidence of unstable conditions such as irregular sloping basement, steep slopes and high gravitational gradients, strong stress regime, intense fracturing, extensively hydrother- Fig. 19. Map showing the potential areas of instability based on the combination of unstable conditions including: slope, alteration areas, fractures, seismic data, and present morphology (modified from Diaz-Castellon, 2003).

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A study of infiltration on three sand capillary barriers

by EC Leong

2004, Canadian Geotechnical Journal

The capillary barrier effect was investigated by conducting infiltration tests on three soil columns of fine sand over medium sand, medium sand over gravelly sand, and fine sand over gravelly sand. The barrier effect was verified in the... more

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Triggering conditions and mobility of debris flows associated to complex earthflows

by Jean-philippe Malet

2005, Geomorphology

Landslides on black marl slopes of the French Alps are, in most cases, complex catastrophic failures in which the initial structural slides transform into slow-moving earthflows. Under specific hydrological conditions, these earthflows... more

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Landslide initiation in saprolite and colluvium in southern Brazil: Field and laboratory observations

by Willy Lacerda

2007, Geomorphology

The weathering of granitic and gneissic rocks in tropical regions can reach depths of more than 100 m. In southeast Brazil there are situations where landslide initiation depends on the fluctuation of the groundwater level, on the impact... more

These facts are also reported by Fookes et al. (2000), who, commenting on their geomorphological model for hot wet climate, state that the depth of weathering can be many tens of metres, controlled by the distribution of discontinuities in the rock mass. Also, duricrusts are for- med where groundwater concentrates soluble weathering products (Thomas, 1994). Deeply weathered colluvial soils are often difficult to distinguish from in-situ weathe- ring profiles. Laterites are hard subsurface deposit of oxides of aluminum and iron found in tropical soils where the water table fluctuates with seasonal changes in preci- pitation. This process is called “‘laterization”, and when Fig. 1. The “Serra do Mar” region in Brazil. Ab’ Saber (2003) describes the main characteristics of saprolitic and colluvial soils in Southeastern Brazil:

Fig. 2. Formation or colluvium under a rock outcrop (Fookes, 1997).

Fig. 3. Normalized shear stress vs. displacement of residual (left column) and colluvial soils (right column) (Lacerda and Silveira, 1992).

Fig. 4. Typical variation of piezometric levels in residual soils from gneissic—granitic rocks (Lacerda, 1989)

Fig. 6. Definition of pore pressures Ug, Umax and up

Fig. 8. Yield surface of granitic residual soil.

Fig. 9. U, vs. number of cycles to failure.

Fig. 10. Shallow slide due to loss of suction.

Fig. 11. The Santa Genoveva Nursing Home accident: cross section (Barros et al., 1989). completely saturate the 5 m thick colluvium, composed of more than 30% of rock blocks, some of which more than 4 m in diameter, at a 20° inclination. Sometimes geologists in Brazil use the term “talus/colluvium” to distinguish this particular type of colluvium. This amount of rain was divided in three distinct periods by Santos Barros and Brandao (1992); the accumulated rainfall in the last 3 days was 441 mm, with a peak hourly rainfall of 56 mm/hr in February 22, the day of the debris flow. The slab fell from elevation 525 m and was broken on its first impact on the rock scarp just below, at elevation 400 m; the resulting flying blocks (with an energy that cut some high trees in half) impacted the colluvium at elevation 280 m. Figs. 11

Fig. 12. The Santa Genoveva Nursing Home accident: plan view (Barros et al., 1989).

Fig. 13. Aspect of the nursing home | day after the landslide (Photo by Ary Maciel). and 12 show the section and the plan view of the site. The colluvium liq declivity of the s home, ate evation in an almost paral which usually carried superficial water storms. T he two uefied, and ran until a 180 m. Another debris s nursing home, indicated by the black arro seen in Fi accident a break in the ope, in which there was a nursing ide occurred el paleo thalweg, also saturated, and flows during debris flows joined exactly at the w, as can be g. 12. About 53 people were killed in this one. Figs. 13 and 14 show res pectively the Saprolitic soils in the saturated state are often sepa- rated from the upper colluvial cover by a clayey soil, which can be an old mature residual soil that was covered by the colluvium, or a layer with a concentration of elluviated, fine material from the more impermeable colluvium. In this case, the aquifer of the saprolitic soil can be confined, and semi-artesian or even artesian pore pressures can develop in the lower stratum if the rate of infiltration from higher elevations is high. The example of a slide due to this cause is that of Urubu’s Hill in the city of Rio de Janeiro, which occurred on April 8, 1966. Several previous slides had been reported by the

Fig. 15. Large fallen rock block, with more than 10 m in maximum dimension, in a forested slope (Tijuca forest, Rio de Janeiro). Photo by the Author. residents of the densely occupied area. These slides happened 30 years before, and it is logical to assume that even when the area was uninhabited they may have also occurred, but had been unnoticed. The slide involved the

Fig. 14. Aspect of the region washed through by the landslide. Gneissic rock appears where a 5 m thick colluvium/talus existed. At right and far lef note the intact colluvium (Photo by the Author in 1989).

Fig. 16. Urubu’s Hill plan (adapted from Moreira, 1974). subsidence of the upper part of the slope, forming a scarp with 4.5 m, cutting a house in half. At the foot of the slope the ground suffered a heave of | m, and the soil and the walls of the houses at this region were cracked. All residents had time to evacuate their houses, because the slide was not sudden. This accident is reported by Moreira (1974). Fig. 16 shows the plan of the site. Rotary drill borings with Denison sampling, piezometers and 4 inclinometers were installed. Sounding S-54, at the side of inclinometer I-4, showed that the subsoil consisted of a colluvium/talus up to 17 m thick resting on a clayey

Fig. 17. Cross section A—A’ showing failure surface and piezometric lines (adapted from Moreira, 1974)

Fig. 18. 25-day accumulated rainfall in the summer of 1966. artesianism. Local artesianism can initiate landslides in upper colluvial layers, as the Author has seen in Brazil.

Fig. 19. Influence of a hidden spring on the flow pattern of a slope (Lacerda, 1999). The remnants of the of Soberbo Road debris flow in the city of Rio de Janeiro, in 1967, have been extensively studied (Soares et al., 1988; Lacerda and Schilling, 1992). The material deposited after the flow constitutes a colluvium, which moves continuously (in reality, intermittently, according to the elevation of the ground water table). Fig. 20 shows a partial plan of the slide. Section C—Cl along zones C, B and D is shown in Fig. 21. The two upper branches of this complex slide present a colluvial layer on top of the residual soil. The thickness of the colluvium lies between 6 and 10 m. The water table is close to the surface, except at two zones, when it appears actually at the surface. At these points impermeable diabase

Fig. 20. Map of the Soberbo Road slide (Lacerda and Schilling, 1992).

Fig. 21. Section along the Soberbo Road Landslide showing two families of diabase dykes (Lacerda, 1999).

Fig. 22. Displacements measured by inclinometers in the region between the diabase dykes (Lacerda and Avelar, 2003) the deepest piezometer just at the upper contact with the dyke rising more than | m above the ground elevation. Of course, the ground was very wet, with rivulets of water springing at the surface. more impermeable dikes influences the flow lines. Thus, artesian pressures can be observed just before the dykes, as shown. In the Soberbo Road case artesianism was indeed observed, the water level of

Fig. 23. Flume set up (Lacerda and Avelar, 2003). Fig. 23 shows the Flume set up (Lacerda and Avelar, 2003). Water reservoirs at the top and bottom of the box are maintained with a constant water level, and observations are made during the steady-state seepage pattern established. Electric and standpipe piezometers

Fig. 25. Slide of the soil mass above the barrier (dyke) (Lacerda and Avelar, 2003). Fig. 24. Sequence of elevation of the water level inside the flume (Lacerda and Avelar, 2003).

Average from several classification tests in each layer (Moreira, 1974) Table 1

Shear strength tests (Moreira, 1974) *DST — direct shear test; submerged; **CIU — isotropically consolidated triaxial test, saturated. Table 2

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Uplift response of buried pipelines in saturated sand deposit under earthquake loading

by Reza Saeedzadeh and

2011, Soil Dynamics and Earthquake Engineering

Pipelines buried in saturated sand deposits, during earthquake loading could damage from resulting uplift due to excess pore water pressure generation. Several studies have been made to better understand the uplift mechanism and evaluate... more

Fig. 9. Maximum shear forces acting on pipe in different density ratios.

Fig. 12. Effect of water table on uplift, Sine Acceleration. Fig. 11. Effect of pipe burial depth on uplift.

Fig. 10. Effect of pipe diameter on uplift.

Fig. 13. Effect of water table on uplift, Naghan Earthquake.

The finite element (FE) code of PLAXIS [27] was used in this study. The software could simulate two dimensional problems under axi-symmetric or plane strain conditions. It also has the capability of dynamic analysis of saturated soil deposits. The soil behavior is simulated by the hardening soil model (HS) as shown in Fig. 1. It is an advanced model for simulating the behavior of different types of soil, both soft soils and stiff soils [28]. Since this model uses the theory of plasticity rather than the theory of elasticity, includes soil dilatancy and introduces a yield cap, super- sedes the well-known hyperbolic model of Duncan and Chang [29]. Fig. 1. Stress—strain relationship of the hardening soil model (HS), [27].

Fig. 4. Experimental and two FEM results of excess pore water pressure, top of pipe. Fig. 3. Experimental and two FEM results of uplift of pipe.

Fig. 5. Experimental and two FEM results of excess pore water pressure, bottom of pipe.

Analyses to evaluate soil dilatancy angle effect. Table 2

Fig. 7. Effect of density ratio of soil on pipe uplift. Fig. 6. Effect of dilatancy angle of soil on pipe uplift.

Fig. 8. Maximum bending moments acting on pipe in different density ratios.

Fig. 14. Effect of water table on uplift, Tabas Earthquake.

Fig. 15. Effect of saturated soil layer thickness on uplift. diameter. In Fig. 14, which shows the results of Tabas earthquake, this phenomenon is more obvious. Presence of this point (for burial depth and pipe diameter of 3 m) in all graphs indicates this fact that if the water table lowering is one of the options of remedial measures against pipeline floatation or prevention of liquefaction occurrence among several other options, there is an optimal height at which the most advantages of lowering the water table can be achieved. Although for a pipeline of hundreds of kilometers length, lowering the water table is not a suitable option, but when this pipeline passes by other structures (e.g. near a refinery) or with a larger diameter, as a tunnel in urban areas, this option can also be taken into account.

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Non-conductive heat transfer associated with frozen soils

by Larry D Hinzman

2001, Global and Planetary Change

The assertion that pure conductive heat transfer always dominates in cold climates is at odds with decades of research in soil physics which clearly demonstrate that non-conductive heat transfer by water and water vapor are significant,... more

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Modeling uncertainty in stability analysis for design of embankment dams on difficult foundations

by Azm S Al-Homoud

2004, Engineering Geology

A probabilistic 3-D slope stability analysis model (PTDSSAM) is developed to evaluate the stability of embankment dams and their foundations under conditions of staged construction taking into consideration uncertainty, spatial... more

Fig. 2. Typical cross section in Dam Body and Foundation at Ch. 1480. To conclude on geotechnical properties, Tables 3 and 4 summarize the principal design parameters for the foundation and dam material. This includes den- sity, effective cohesion and angle of friction, perme- ability, total friction and angle of friction, compress- 1 t d t ibility and consolidation coefficients, etc. Table 5 lists he values for the pore pressure parameter for the ifferent materials used in the dam construction and he foundation material as used by Sir Alexander Gibb and Partners (1988) in the design of the dam prior to any measurement of loading-induced pore pressure a uring staged construction of the dam.

Fig. 1. The main zones of typical cross section of Karameh Dam Project (modified after Sir Alexander Gibb and Partners, 1993), of-the-art testing equipment and following strictly the related ASTM standards.

Fig. 3. Detailed cross section in Dam Body at Ch. 1480.

Fig. 4. Exponent parameter (m) and normally consolidated strength value (S) for the normalized undrained strength relations of foundation material along Karameh Dam.

Fig. 5. The critical failure surface for USA and ESA stability analysis methods for cross-section at Ch. 1480 in Karameh Dam (modified after Sir Alexander Gibb and Partners, 1993). Fig. 4 shows the normalised undrained strength relations developed here for the foundation material along Karameh dam. Three chainages Ch 800, 1480, and 1850 are analyzed using PCSTABLSM program at different stages of construction. In ESA method, drained soil properties are used. The following effective shear strength parameters are used in the analysis based on

Fig. 6. Variation of safety factors for stability of critical failure surfaces at final stages of fill at different dam sections along Karameh Dam (obtained using PCSTABLSM Program). USA safety factors are evaluated based on lower and upper bound of undrained shear strength S, in fill and foundation material. Typical safety factor values for critical failure surface vs. height of fill are obtained for the main dam sections. Results are obtained based on monitored loading-induced pore water pressure. The evaluated safety factors are compared with Sir Alexander Gibb and Partners (1997) for the final stage of construction at Ch 800 and 1480. The variation of safety factors of critical failure surfaces at final stage of fill at different dam sections along Karameh dam using undrained Typical results of analysis are presented for the dam section at Ch 1480. Fig. 5 shows the critical failure surface for USA and ESA for the cross-section at Ch 1480. The evaluated safety factors tend to decrease with increasing the height of fill. The upper bound of computed safety factors are those obtained using ESA based on field pore water pressure data, while the lower bound are those computed using ESA with B=0.9 in foundation. These results indicate that the dam is stable during construction, and the actual rate of loading is controlled.

Fig. 7. View of Earth slope and failure mass; and cross section within failure mass. The uncertainty in driving moment oy, is neglected in this study, as the unit weight of the soil responsible for driving moment is assumed to be deterministic. Those statistical parameters depend on the spatial average of the shear strength properties. Spatial aver- age of shear strength differs from point value of shear strength. According to Al-Homoud (1985), if S and s are the point and spatial average values of the un- drained shear strength, S is the standard deviation, and §=r.(L)S, see Fig. 7, where L=total arc length: L =

Fig. 8. Simplified flow chart for PTDSSAM analysis.

Fig. 9. Effect of c.o.v. on probability of slope failure Pf, for the main sections along Karameh Dam at different stages of construction fill using USA method incorporated in PTDSSAM Program.

Fig. 10. Effect of total slope width B, on probability of slope failure Pf, for the main sections along Karameh Dam at different stages ot construction fill using USA method incorporated in PTDSSAM Program.

c’' =effective cohesion, ¢/ =effective angle of friction, s,=un- drained shear strength. *Lower bound derived from consolidation undrained (CU) triaxial test. ** Upper bound.

Thickness and elevation of main units (Sir Alexander Gibb and Partners, 1993) SAMRA formation

The input and output parameters of PTDSSAM analysis Table 6

Correlation between S, and ¢@ within same material Correction factors for sy and @

Scale of fluctuation of undrained shear strength in x, y, and z direction Coefficient of variation for undrained shear strength and angle of friction for different materials

Coefficient of variation of correction factor of s,

Coefficient of variation of correction factor of d

Variation of safety factor against foundation failure in 2-D and 3-D, total probability of failure, (ps), reliability index, (f), and critical failure width, (b,) with the correction factor for modeling error, (Vj), coefficient of variation of N;, Q(N;), progressive failure correction factor (N,.), (QUN,,)), total width of slope (B) scale of fluctuation of s, in z direction, 2,(s,), and 7, = ,(sy) for Ch 1480 based on probabilistic approach using PTDSS program (Al-Homoud, 1985) and undrained strength analysis (USA) method

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Erosion Processes in Gullies Modified by Establishing Grass Hedges

by Eddy Langendoen

2000

Concentrated flow can cause gully formation on sloping lands and in riparian zones of floodplains adjacent to incising stream channels. Current practice for riparian gully control involves blocking the gully with an earthen embankment and... more

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Laboratory tests for measuring the time-lag of fully grouted piezometers

by L. Simeoni

2012, Journal of Hydrology

Time-lag is the most important parameter describing the conformance of a piezometer because it represents the time that it takes for the instrument to reach equilibrium when there is a pore water pressure change in the soil. The use of... more

Fig. 6. Tests imposing an increment of the water pressure at the bottom of the sample: relation between U;op and time ft. (a) Sample S4-03; (b) sample S8-03

This article appeared in a journal published by Elsevier. The attached copy is furnished to the author for internal non-commercial research and education use, including for instruction at the authors institution and sharing with colleagues. Provided for non-commercial research and education use. Not for reproduction, distribution or commercial use.

Fig. 1. Piezometer installation: (a) conventional installation; (b) grouted-in installation.

Fig. 2. Sketch of the testing apparatus.

Fig. 5. Increments of water pressure at the top of samples S8-01 and S8-02 for an imposed increment of 100 kPa at the bottom. Test B: from 100 kPa to 200 kPa, test C from 200 kPa to 300 kPa. Fig. 4. Water pressure response at the top of the samples S2, S4 and S8 for an imposed pressure change at the bottom. (a) Pressure increment from 100 kPa to 200 kPa (test B); (b) pressure decrement from 200 kPa to 100 kPa (test G).

Fig. 3. Water pressure response at the top of the sample S8-01 for an imposed pressure change at the bottom. (a) Pressure increment from 100 kPa to 200 kPa (test B); (b) pressure decrement from 200 kPa to 100 kPa (test G).

Fig. 8. Relation between to; and the average water pressure value at the bottom of the sample. Closed and open symbols refer to upflow (uw) and downflow (dw) tests. Fig. 7. Sample S8-03: relation between Ujop and time t for tests uw (upflow) imposing an increment of the water pressure (A, B, C, D) and tests dw (downflow) imposing a decrement of the water pressure (E, F, G, H).

the change in water pressure by a coefficient a, or equivalently with the change in piezometric head by the coefficient b = ayw:

Fig. 9. Relation between tos and height for samples S4-03 and S8-03. Closed and open symbols refer to upflow (uw) and downflow (dw) tests. According to Eq. (7) tgs should depend only on the height of the sample, and not also on the value of the water pressure as was ob- served during the tests (Fig. 8). Fig. 9 shows the relation between tg; and H for the samples S4-03 and S8-03. In order to reduce the differences in the saturated hydraulic conductivity the two sam- ples were prepared with the same mixture (tremieing group 3) and were tested at a similar maturation time (50 and 48 days). It was seen that tos generally increased with H according to Eq. (7), but different values were calculated for the same height. In fact, tos ranged between 0.56 min and 1.61 min for sample S4-03 and between 1.09 min and 6.54 min for sample S8-03.

Fig. 10. Measured and computed degree of excess water pressure at the top, Urop, for test B on sample S8-03.

Fig. 11. Experimental and modelled data for test B on sample S8-03. Typical relationships between the volumetric water content versus the matric suction (retention curve) and the hydraulic con- ductivity versus the volumetric water content are shown in Fig. 12a and b. Fig. 12a shows that the relationships between vol- umetric water content and matric suction usually exhibits hyster- esis. For a given value of matric suction the volumetric water content during drying is greater than the volumetric water content in wetting. The hydraulic conductivity (Fig. 12b) is generally ob- served to be uniquely related to the volumetric water content (Fredlund and Rahardjo, 1993), independently on the drying or wetting path, and increases sharply when the volumetric water content approaches the value at saturation. Bicalho, 1999 studied the hydraulic behavior of quasi-saturated silty soils having a de- gree of saturation from 0.9 to 1 and in which the air phase was present in the form of occluded bubbles. The Author found that sat- uration was achieved only with positive water pressures and that the hydraulic conductivity increased of an order of magnitude.

Fig. 13. Volumetric water content function assumed for the analyses with SEEP/W. Pressure is in kPa. To explain the dependence of c, on the sample height finite ele- ment analyses were performed to study the pore water pressure evolution by using the program SEEP/W. Steps 50-100 kPa, 100- 200 kPa and 200-300 kPa were analysed on samples of 2 cm, 4cm and 8 cm heights. The hydraulic conductivity and volumetric water content relationships assumed for the grout are shown in Figs. 13 and 14 versus the pore water pressure. An hydraulic con- ductivity of 2.65 x 10~* m/s was assumed at saturation. During the analyses the flux through the lateral and top surfaces was assumed

Fig. 12. Typical relationships between suction (u,-u,,) versus (a) volumetric water content, and (b) relative hydraulic conductivity k/k,, where k, is the hydraulic conductivity at saturation.

Fig. 15. Geometry objects and boundary conditions: (a) membrane tightly adherent to the soil sample; (b) thin interspace in between membrane and soil.

Fig. 14. Hydraulic conductivity function assumed for the analyses with SEEP/W. Pressure is in kPa and conductivity is in m/s.

Fig. 16. Numerical analysis for bottom pressure from 200 kPa to 300 kPa. Total head at 0.3s, 10.2s and 41.0s in a 8cm sample with (wider line) or without (thinner line) the boundary water interspace. Values of cy fitting the results of SEEP/W analyses with sample-membrane interface. Table 4

List of samples. Table 2 demonstrated that a difference greater than 80 kPa between the cell pressure and the water pressure at the bottom of the sample was needed to prevent any seepage through the sample. although the procedure for the sample preparation was the same in the three tremieing groups. Fig. 5 shows the increments of water pressure at the top of samples S8-01 and S8-02 when an increment of 100 kPa was imposed at the bottom. Two tests are illustrated for both samples: test B refers to the increment from 100kPa to 200 kPa; test C refers to the increment from 200 kPa to 300 kPa. Even though the samples were subjected to the same hydraulic gradient, the response time is shorter when the imposed pressure is greater.

Values of cy, fitting the results of SEEP/W analyses without sample-membrane interface. Table 3

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Application of Groundwater Modeling in Landslide Assessment

by Binaya Paudel

Monitoring and analysis of hydrological parameters are crucial to understand the underlying causes and mechanisms of a landslide. This handout analyses hydrological parameters of an existing creeping landslide site in western Japan, as an... more

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Incorporating bank-toe erosion by hydraulic shear into a bank-stability model: Missouri River, Eastern Montana

by Eddy Langendoen

Bank-stability concerns along the Missouri River, eastern Montana are heightened by a simulated change in flow releases from Fort Peck Dam to improve habitat conditions for Pallid Sturgeon. The effects of the simulated flow releases on... more

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Causes of catastrophic failure of Tam Pokhari moraine dam in the Mt. Everest region

by Rabindra Osti

2011, Natural Hazards

The moraine dam of the Tam Pokhari glacial lake breached on 3 September 1998 and caused a catastrophic flood in the downstream areas. To learn from the event, a field survey was conducted. The survey team found that a landslide, which is... more

round 6:00 with its peak around mid night (Bollasina et al. 2002; Ueno et al. 2001). According to Ueno et al. (2001), 79% of the annual rainfall in the region was observed in he monsoon season during the period from 1995 to 1999 (observation station located at 36°43’E, 27°49’N, 3,833 m asl). The region’s daily low and high temperatures for the same yeriod fluctuated between —11 and 11°C, and January and August were the coldest and warmest months, respectively. ee ee ee a a ee i rr 1), a ee ee ee a ee ee

Fig. 2 Geometrical feature of Tam Pokhari moraine dam Figure 3 shows the upstream facing pictures of the existing lake. As shown in Fig. 3a, the lake receives melt water from the existing glacier (the Sabai Tsho glacier), which snout is located about 150 m upslope to the north from the lake head. In the space between the glacier tongue and the lake is a steep and narrow cliff composed of massive gneiss. The present outlet, the origin of the Inkhu Khola, lies at the end of the lake and exactly at

Fig. 3. Lake level down-draw (50 m), damaged moraine dam and debris plunged into the lake (camera facing upstream (a) and downstream (b))

Fig. 4 Difference in sediment composition in the /eft and right banks of dam breached section

Fig. 5 Grain size distribution of deposited sediment mass at the lake mouth

layering structure is also lacking. An inclined thin layer (dip amount 45° due northeast) of clayey (crushed) materials marks the contact between the edges of underlying inclined sediment layers and the overlying unconsolidated moraine materials. The clayey materials consist of a very few gravels with their longer axis aligned parallel to the 45-degree inclination of the clay layer. It is to note that the layering structure of dam extended up to the upstream face of the dam in the past, but the structure was damaged along with the downward movement of the upstream-facing slope (a mass slide without detachment from the dam), whose stability became critical after glacier melt. These features of the upstream- facing part of the dam suggest the previous movement of sediment mass within the dam and further indicate the fact that the contact plane could have been the share plane of the landslide.

Fig. 7 Observed monthly rainfall at Chaurikharka Hydrological station

Fig. 8 Observed 6 days daily rainfall at Chepuwa and Chaurikharka (after Ueno et al. 2001)

Fig. 9 Sabai Tsho Glacier before glacial lake outburst (Jeft—Sponsortrek 2008; right—Photo by Michael Durham dated October 2005 with permission) \Q m) and the difference in surface area. According to local people, the overflow from the ake continued for several hours, while the maximum flow lasted for about half an hour. Snow avalanches are often considered to he region. Some researchers (e.g. Vuichard fact that the glacial lake had breached in egion. In the case of Tam Pokhari Lake, a ause of dam failure (Dwivedi et al. 2000) ecessarily on the same day of the dam breac have caused many historical GLOF events in and Zimmermann 1987) have also pointed out the past due to snow or ice avalanches in the snow avalanche was considered as the sole . A clear evidence of snow avalanche (not h) is also the snout of the existing Sabai Tsho lacier, located in the northern part of the lake, which has shifted about 150 m upslope than | was in the past (Figs. 3b, 9). However, if allen into the lake from a glacier, it would a huge mass of snow or Ice avalanche had not have been able to cause much impact ecause it would have rather formed a pile of snow or ice mass at the glacier foot where the ike is narrow and shallow. It is also to note that the lake shape does not permit any wave riginating from the lake-head to propagate in a single direction; such a wave usua educes its strength due to to-and-fro motion between east and west banks. Therefore, it ossible to say that a snow or ice avalanche plunged into the lake, displaced the lake water nd generated surges that propagated up to the dam and raised the water level for a whi ut this cannot be the direct cause of the dam breach. Lacking any evidence of impact on he eastern and western sides of the lake also does not permit to justify the occurrence uge surge induced by a snow or ice avalanche. Additionally, the left and right sides of t ike are steep (more than 60°) and stable mountains (Fig. 3), where a huge snow mass may ot be able to retain for a long period; therefore, chances of snow avalanche generation rom either side slope is also minimum. ly iS e, of he Similarly, there were also no signs of the ccurrence of rock avalanches from the side slopes around the lake, which indicates that he dam breach was not caused by a rock avalanche, either. The mnaraine dam lacratinn lake chane enrrainnding tannagranhy and antlet canditian

the dam whose stability had been already in a critical state due to the increase in pore water pressure and decrease in height and width. A huge mass of slid moraine materials still remain deposited on the foot of the moraine dam as described. The sediment mass could not have been fully flushed except the part at the centreline of the flow. iv) The increased discharge at the outlet due to the increased outlet capacity further helped the dam failure process by rapid erosion. As a result, the dam breach occurred within 9 chort time inducino a hnoe water flix which micht have lacted far ahant 20 min The stability analysis based on the SLOP/W module in the GeoStudio model results in the landslide with a slip circle radius of 170 m (Fig. 10) and a safety factor of 0.85. This is quite reasonable for a case of failure. The detail of acting forces (in KN) and momentum calculated by using the Morgenstern-Price Method for slice number 13 (shaded in Fig. 10) as an example is illustrated in Fig. 11. It has been found that the slip circle coincides with pre-existing share plane in the dam. In this result, the change in pore water pressure due to

Table 1 Earthquake events recorded within 300 km distance from the Tam Pokhari Glacial Lake during 15 August to 15 September 1998 (USGS 2009) * Epicentre located within 15 km radius from the Tam Pokhari Glacial Lake

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Mini suction probe for matric suction measurements

by Meilani We

2002, Canadian Geotechnical Journal

A modified triaxial apparatus with mini suction probes was fabricated to study the matric suction along the specimen height during unsaturated triaxial testing. Three mini suction probes were placed at 3/4, 1/2, and 1/4 height of the... more

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Pore water pressure and streambank stability: results from a monitoring site on the Sieve River, Italy

by Alessandro Gargini

1999, Earth Surface Processes and Landforms

To investigate the role of pore water pressures in the stability of a streambank, a series of tensiometers and piezometers was installed in a bank of the Sieve River, Tuscany, Italy. Fluvial entrainment at the bank toe was monitored by... more

Figure 1. Location map for the Sieve River, central Italy. (A) Sieve River course. (B) Location of the study site. Key: 1, monitored bank; 2, gauging station; 3, alluvial deposits

Figure 3. Sketch of the monitored reach. (A) Monitoring station. (B) River reach covered by fluvial entrainment study. Key: 1, Silt and sand; 2, loose gravel; 3, packed gravel with interstitial sand MONITORING INSTRUMENTATION

Figure 4. Results of 16-month monitoring programme. (A) Rainfall, river stage and groundwater level. (B) Trend of the matric suctiot measured at the tensiometers (T1, T2, T3, T4, TS from the shallowest to the deepest).

Figure 5. Record of 14 December 1996 flow event. (A) Rainfall and river stages. (B) Variations in suction at the four tensiometers

Figure 6. Record of 15 February 1997 flow event. (A) Rainfall, river stage and groundwater level. (B) Variations in suction at the five tensiometers

Figure 7. Changes of total head and hydraulic gradient during the 16 months of monitoring. (A) Trend of the total head. (B) Trend of the vertical hydraulic gradients

Figure 8. Geometry of the bank and mechanism of failure. T1-T5, Piezo-tensiometers; GW, piezometer Copyright © 1999 John Wiley & Sons, Ltd.

Figure 9. Trend of the safety factor (Ff) through time compared with the river stage (H) during the monitored period

Figure 10. Bank stability and erosion during the 14 December 1996 flow event. (A) Variation in the factor of safety (F’,) compared with the river stage (H). (B) Variation in the packed and loose gravel stability numbers (NV) compared with the river stage (H)

Figure 11. Variations in the loose and packed gravel stability numbers (N,) during the monitoring period

Figure 12. Repeated bank profiles 1995-1997. Key: 1, silt and sand; 2, loose gravel; 3, packed gravel with interstitial sanc

Table I. Main morphological, sedimentological and hydrological data of the study site at Fornacina

Table II. Details of monitoring instrumentation

Table III. Peak stage (H), estimated mean boundary shear stress (7,) and shear stress on the bank (7;) for the four flow peaks during the period of fluvial entrainment measurements

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Physically-based and distributed approach to analyze rainfall-triggered landslides at watershed scale

by Elisa Arnone

2011, Geomorphology

Landslides are a serious threat to life and property throughout the world. The causes of landslides are various since multiple dynamic processes are involved in driving slope failures. One of these causes is prolonged rainfall, which... more

where p, and py, are soil and water densities (kg/m?); h, is soil thickness (m). This classical model, that relates instability potential to Fig. 1. Flow chart of landslide system in tRIBS model.

Fig. 2. Geometry of the infinite slope with a soil thickness h,, a groundwater level h anc a slope angle a. A unit element b parallel to the horizontal surface is considered. W, T, N and U are the forces acting on a unit slice of soil, described by Eqs. 2-5.

A 25 m resolution digital elevation model (DEM) is available for the area (Fig. 3), and it has been transformed into the irregular mesh used by tRIBS (TIN) by using the hydrographic TIN method developed by Vivoni et al. (2004). This method significantly reduces the number of elevation nodes preserving the distribution of topographic attributes as slope and curvature. Here a percentage of 20% of DEM nodes has been retained in the TIN geometry since this percentage The main purpose of this application is to demonstrate the physical consistency of the proposed methodology and to test the model performance in term of landslide occurrences in response to the climatic forcing. A small catchment located in western Sicily (Italy)

Fig. 5. Time series of relative wetness index (Ry) (a) and factor of safety (FS) for three different soil thickness (H1, H2, H3) (b), cohesion (C1, C2, C3) (c) and friction angle (F1, F2, F3 (d), of a sandy clay loam element with a slope of 26° and for rainfall event R3.

Fig. 6. Factor of safety (FS) as a function of soil thickness (H), cohesion (C) and friction angle (F), for the three rainfall events (R1, R2, R3), two slopes (S1 and S2) and the two soil types (sandy loam ST2 and sandy clay loam ST3). First step of the landslide module is the correct classification of unconditionally and conditionally stable elements. The empirical frequency distribution of slope (Fig. 7a) is evaluated for the entire basin and then split by unconditionally and conditionally stable elements. Since the spatial distribution of soil type is homogeneous with regard to the geotechnical parameters, it is possible to identify a the sensitivity analysis showed how the variability of C and F is not crucial, which instead is a good description of the geomorphology (slope) and the choice of H. In the simulations a value of H= 1.5 m has been selected; this choice was based on typical values of shallow landslides and to keep the hypothesis of the infinite slope approach (longitudinal length much greater than soil thickness) valid.

Fig. 7. Frequency distribution of slope relative to total, unconditionally stable, conditionally stable (a) and landsliding elements (b).

For each time interval, the model evaluates FS values, and flags elements with FS = 1 as landslides. It then calculates the possible path to be followed by the material after its detachment, according to the approach described in Section 3.2. Fig. 8b shows the distribution of landsliding cells at the last simulation interval, together with the path of the landslides. Each head of a landslide corresponds to an element whose factor of safety less than 1. Material comes to a halt as soon as it Finally, the comparison between observed and simulated land- slides is shown in Fig. 10. The historical landslides that occurred in the southeastern part of the basin, along the hillslope between the river and the eastern boundary of the basin, have been correctly

Fig. 10. Comparison between documented events and simulated events. A new landslide component of the distributed hydrological model tRIBS is here presented. The model estimates where and when a failure phenomenon could occur, by simulating most of the hydrological processes present in a hillslope, and by evaluating its stability based on the infinite slope model. The hypothesis of this

Fig. 9. Correlation of landslide with slope and wetness index (a) and groundwater table depth (GW) and wetness index (b).

Values of geotechnical parameters used in the sensitivity analysis and in the model application. Table 3

is the pore-size distribution index, y is the air entry bubbling pressure, f is the decay parameter of conductivity, A, is the saturated anisotropy ratio, A, is the unsaturate isotropy ratio, n is the total porosity, k, is the volumetric head conductivity and C, is the soil heat capacity.

Durations and total volumes of rainfall events considered.

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Modelling the effects of vegetation on stability of slopes

by M. Jaksa

2004

It is well understood that vegetation influences slope stability in two ways: through hydrological effects and mechanical effects. Hydrological effects involve the removal of soil water by evapotranspiration through vegetation, which lead... more

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Impact of tunneling on regional groundwater flow and implications for swelling of clay–sulfate rocks

by Christoph Butscher

2011, Engineering Geology

Tunnels play a key role in many transportation concepts. The swelling of clay-sulfate rocks leads to serious damage to many tunnels crossing such rock, producing great difficulties and high extra costs in tunnel engineering. The swelling... more

Fig. 1. Study area and location of the cross-section of Fig. 2 (arrow).

ig. 2. Top: Geological cross-section of the study area (Figure 3 can serve as legend). The cross-section is the basis for the finite element groundwater models used in this study ottom: Model area, boundary conditions and solution for the hydraulic head field of the most simple model setup (homogeneously distributed hydraulic conductivity, withou unnel and faults). The rectangles mark the detail of the model area which is described in the text and illustrated in Figs. 4-7.

Fig. 5. Hydraulic head field and flow paths towards the tunnel perimeter calculated with the homogeneous model setup with faults. Left: without tunnel; right: with tunnel subsurface feature. At the position of the tunnel, but without the tunnel as structural element (Figure 4 left), the groundwater flow is directed upward, with a pronounced horizontal component. The capture zone of the water passing through the (future) tunnel is a very narrow area originating from the hill in the north of the regarded model detail (in the following and in Figures 4-7, this hill is addressed as the “northern hill”). After the tunnel excavation (Figure 4 right), the capture zone is largely broadened and distributed to the northern hill and to other minor hills. The individual recharge areas on the hills are also broadened. One flow path originates from the valley above the tunnel. This section presents the results of the numerical simulations using the different model setups indicated in Table 1. The distribution of the hydraulic head and flow paths towards the tunnel, visualized by backward tracking of 20 particles from the tunnel perimeter as starting points to the ground surface (or model boundary), are used to

Fig. 4. Hydraulic head field and flow paths towards the tunnel perimeter calculated with the homogeneous model setup without faults. Left: without tunnel; right: with tunnel as subsurface feature.

ig. 6. Hydraulic head field and flow paths towards the tunnel perimeter calculated with the aquifer-aquitard model setup. Left: without tunnel; right: with tunnel as subsurfac ‘ature.

‘ig. 7. Hydraulic head field and flow paths towards the tunnel perimeter calculated with the hydrostratigraphic model setup. Left: without tunnel; right: with tunnel as subsurface eature.

Overview of considered subsurface features and corresponding model setups. Table 1

Fig. 3. Hydrostratigraphic units of the study area and respective hydraulic properties attributed to the finite element groundwater models. Lithostratigraphic column adapted from Bitterli-Brunner and Fischer (1988).

Flow rates in liters per day per m tunnel section (I/d per m) into the tunnel perimeter calculated for the different model setups (c.f. Table 1) before and after the tunnel excavation Note: The generally lower flow rates in the aquifer-aquitard and hydrostratigraphic models compared to those of the homogeneous models are due to the lower hydraulic conductivities at the position of the tunnel in these models.

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Comparative analysis of contributing parameters for rainfall-triggered landslides in the Lesser Himalaya of Nepal

by Santosh Dhakal

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In the Himalaya, people live in widely spread settlements and suffer more from landslides than from any other type of natural disaster. The intense summer monsoons are the main factor in triggering landslides. However, the relations... more

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Grass Hedge Effects on Gully Hydraulics and Erosion

by Eddy Langendoen

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Concentrated flow can cause gully formation on sloping lands and in riparian zones adjacent to incising stream channels. Current practice for riparian gully control involves blocking the gully with a structure comprised of an earthen... more

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Spatial and temporal variability of pore-water pressures in residual soil slopes in a tropical climate

by EC Leong

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It is critical to understand and quantify the temporal and spatial variability in hillslope hydrological data in order to advance hillslope hydrological studies, evaluate distributed parameter hydrological models, analyse variability in... more

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Monitoring and modelling of pore water pressure changes and riverbank stability during flow events

by Alessandro Gargini

2004, Earth Surface Processes and Landforms

Pore water pressures (positive and negative) were monitored for four years (1996–1999) using a series of tensiometer-piezometers at increasing depths in a riverbank of the Sieve River, Tuscany (central Italy), with the overall objective... more

Pore water pressures (positive and negative) were monitored for four years (1996–1999) using a series of tensiometer-piezometers at increasing depths in a riverbank of the Sieve River, Tuscany (central Italy), with the overall objective of investigating pore pressure changes in response to ﬂow events and their effects on bank stability.The saturated/unsaturated ﬂow was modelled using a ﬁnite element seepage analysis, for the main ﬂow events occurring during the four-year monitoring period. Modelling results were validated by comparing measured with computed pore water pressure values for a series of representative events. Riverbank stability analysis was conducted by applying the limit equilibrium method (Morgenstern-Price), using pore water pressure distributions obtained by the seepage analysis.The simulation of the 14 December 1996 event, during which a bank failure occurred, is reported in detail to illustrate the relations between the water table and river stage during the various phases of the hydrograph and their effects on bank stability. The simulation, according to monitored data, shows that the failure occurred three hours after the peak stage, during the inversion of ﬂow (from the bank towards the river). A relatively limited development of positive pore pressures, reducing the effective stress and annulling the shear strength term due to the matric suction, and the sudden loss of the conﬁning pressure of the river during the initial drawdown were responsible for triggering the mass failure.Results deriving from the seepage and stability analysis of nine selected ﬂow events were then used to investigate the role of the ﬂow event characteristics (in terms of peak stages and hydrograph characteristics) and of changes in bank geometry. Besides the peak river stage, which mainly controls the occurrence of conditions of instability, an important role is played by the hydrograph characteristics, in particular by the presence of one or more minor peaks in the river stage preceding the main one. Copyright © 2004 John Wiley & Sons, Ltd.

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Propagation of seismic waves through liquefied soils

by Boris Jeremić

2010

To predict the earthquake response of saturated porous media it is essential to correctly simulate the generation, redistribution, and dissipation of excess pore water pressure during and after earthquake shaking. To this end, a reliable... more

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Pore Water Pressure

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